PROCESS DESIGN MANUAL
FOR
UPGRADING EXISTING
WASTEWATER TREATMENT PLANTS
U. S. ENVIRONMENTAL PROTECTION AGENCY
Technology Transfer
October 1974
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ACKNOWLEDGMENTS
The original edition of this Design Manual was published in October 1971. This first revision
to the original text was prepared by Metcalf & Eddy, Inc.; the original manual was prepared
by Roy F. Weston, Inc. Major EPA contributors and reviewers were Richard C. Brenner and
John M. Smith of the U. S. EPA National Environmental Research Center, Cincinnati, Ohio,
and Denis Lussier of the Office of Technology Transfer. Revisions to the text were under
the direction of Donald E. Schwinn and Richard B. Gassett of Metcalf & Eddy, Inc.
rp ,TT
NOTICE
The mention of trade names or commercial products in this publication is for illustration
purposes and does not constitute endorsement of recommendation for use by the
U. S. Environmental Protection Agency.
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ABSTRACT
The main purposes of this manual are to examine situations that necessitate upgrading of
existing municipal wastewater treatment plants and to discuss and evaluate the corrective
actions that are required to upgrade these existing plants. Upgrading to overcome organic
and hydraulic overloadings and/or to meet more stringent treatment requirements is
considered.
The manual emphasizes that operational improvement and modifications to existing unit
operations be considered as the logical initial approach to upgrading existing treatment
plants, before major expansion of existing facilities is implemented.
Because of the numerous alternatives available for upgrading an existing treatment plant, it
is necessary to understand thoroughly the fundamentals of the various unit operations
commonly used in municipal wastewater treatment plants. Therefore, this manual examines
in depth the capabilities, limitations, and interrelationships of the various unit processes.
The manual also examines hypothetical situations requiring upgrading of unit operations
and describes "order of magnitude" costs associated with the upgrading of various unit
operations.
One chapter of the manual presents case histories of upgrading of existing wastewater
*"* nent plants to illustrate the approaches actually used in these circumstances.
111
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CONTENTS
Chapter Page
ACKNOWLEDGMENTS ii
ABSTRACT "»
CONTENTS v
LIST OF FIGURES ix
LIST OF TABLES xiii
FOREWORD xvii
1 INTRODUCTION 1-1
2 INVESTIGATIVE APPROACH
2.1 Purpose of Upgrading Existing Wastewater Treatment Plants 2-1
2.2 Identification of Existing Problem Areas 2-1
2.3 Upgrading to Meet More Stringent Treatment Requirements 2-2
2.4 Upgrading to Relieve Hydraulic and Organic Overloads 2-2
2.5 Upgrading to Improve Plant Design and Operation 2-8
2.6 Consideration of Applicable Upgrading Techniques 2-10
2.7 References 2-11
v-
3 FLOW EQUALIZA,
3.1 Introduction and Concept 3-1
3.2 Benefits of Dry Weather Flow Equalization 3-4
3.3 Determination of Equalization Requirements 3-6
3.4 Costs 3-15
3.5 Performance and Case Histories 3-16
3.6 References 3-21
4 TECHNIQUES FOR UPGRADING TRICKLING FILTER PLANTS
4.1 General 4-1
4.2 Trickling Filter Processes 4-1
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CONTENTS - Continued
Chapter Page
4.3 Trickling Filter Design Considerations 4-3
4.4 Trickling Filter Upgrading Considerations 4-12
4.5 Trickling Filter Upgrading Techniques and Design Basis 4-17
4.6 References 4-48
5 TECHNIQUES FOR UPGRADING ACTIVATED SLUDGE PLANTS
5.1 General 5-1
5.2 Activated Sludge Processes 5-1
5.3 Activated Sludge Design Considerations 5-19
5.4 Pilot Studies 5-28
5.5 Activated Sludge Upgrading Techniques and Design Bases 5-36
5.6 References 5-55
6 CLARIFICATION AND CHEMICAL TREATMENT
6.1 Advantages of Upgrading Clarifiers 6-1
6.2 Process Design of Clarifiers 6-2
6.3 Physical Upgrading of Clarifiers 6-5
6.4 Chemical Treatment 6-14
6.5 References 6-22
7 EFFLUENT POLISHING TECHNIQUES
7.1 General 7-1
7.2 Polishing Lagoons 7-1
7.3 Microscreening 7-7
7.4 Filtration 7-12
7.5 Activated Carbon Adsorption 7-23
7.6 References 7-36
8 PREAERATION AND POSTAERATION PRACTICES
8.1 Preaeration 8-1
8.2 Postaeration 8-4
8.3 References 8-13
VI
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CONTENTS - Continued
Chapter Page
9 DISINFECTION AND ODOR CONTROL
9.1 General 9-1
9.2 Disinfection 9-1
9.3 Odor Control 9-10
9.4 Other Uses of Chlorine 9-15
9.5 References 9-16
10 SLUDGE THICKENING
10.1 Sludge Treatment 10-1
10.2 General Sludge Thickening Considerations 10-6
10.3 Gravity Thickening 10-7
10.4 Air Flotation 10-12
10.5 Centrifugation 10-19
10.6 References 10-25
11 SLUDGE STABILIZATION
11.1 General 11-1
11.2 Anaerobic Digestion 11-1
11.3 Aerobic Digestion 11-17
11.4 Heat Treatment of Sludge 11-23
11.5 Lime Stabilization of Sludge 11-27
11.6 References 11-28
12 SLUDGE DEWATERING
12.1 General 12-1
12.2 Vacuum Filtration 12-1
12.3 Drying Beds 12-13
12.4 Centrifugation 12-17
12.5 Filter Presses 12-24
12.6 References 12-25
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CONTENTS - Continued
Chapter Page
13 CASE HISTORIES OF TREATMENT PLANT UPGRADING
13.1 Case History No. 1 Use of Roughing Filter to Upgrade
an Existing Low-Rate Trickling Filter Plant 13-1
13.2 Case History No. 2 Upgrading an Existing High-Rate
Trickling Filter by Conversion to a Super-Rate
Filter System 13-1
13.3 Case History No. 3 Upgrading Using Polyelectrolyte
Addition Before the Primary Clarifier 13-5
13.4 Case History No. 4 Upgrading a Trickling Filter
Plant by Adding Activated Sludge Treatment and
Pre-and Post-Chlorination 13-9
13.5 Case History No. 5 Upgrading a Primary Treatment
Plant to Provide Tertiary Treatment 13-11
13.6 Case History No. 6 Upgrading a Trickling Filter Plant
in Stages to an Activated Sludge Plant with Roughing
Filters 13-19
13.7 Case History No. 7 Upgrading by Optimization of a"-
Aeration Tank-Clarifier Relationship 13-25
13.8 Case History No. 8 Upgrading by Optimization of
Aeration Tank-Clarifier Relationship 13-28
13.9 Case History No. 9 Upgrading a Modified Aeration
System for Nutrient Removal 13-30
13.10 References 13-36
APPENDIX A -METRIC CON VERSION CHART A-l
APPENDIX B - WORD ABBREVIATIONS B-l
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LIST OF FIGURES
Figure No. Page
3-1 Schematic Flow Diagrams of Equalization Facilities 3-3
3-2 Raw Wastewater Flow and BOD Variation before Equalization 3-8
3-3 Hydrograph for Typical Diurnal Flow 3-9
3-4 Raw Wastewater Flow and BOD Variation after Equalization 3-12
3-5 Earthen Equalization Basin 3-14
3-6 Effect of Tannery Flow Equalization 3-18
3-7 Walled Lake-Novi Wastewater Treatment: Plant 3-20
4-1 Comparison of Trickling Filter Operating Data with NRC
Equation 4-7
4-2 Common Flow Diagrams for Single and Two-Stage High-Rate
Trickling Filters 4-14
4-3 Upgrading a Single-Stage Low-Rate TricMing Filter by
Improving Distribution 4-19
4-4 Modifying a Single-Stage Trickling Filter to a Two-Stage
Filtration System 4-22
4-5 Upgrading a Single Stage Trickling Filteir to a Two-Stage
Biological Filtration/Activated Sludge System 4-24
4-6 Upgrading a Trickling Filter System to rtovide Year Round
Nitrification 4-31
4-7 Upgrading a Single-Stage Trickling Filter to a Two-Stage
Filtration System to Provide Nitrification 4-32
4-8 Upgrading a Single-Stage Trickling Filter to a Two-Stage
Biological System to Provide Nitrification 4-35
4-9 Upgrading a Standard Rate Trickling Filter System with
Chemical Addition to Provide Phosplhorus Removal 4-39
4-10 Upgrading a Trickling Filter System Using Chemical
Addition for Phosphorus Removal 4-44
4-11 Upgrading a Trickling Filter System to Provide Phosphorus
Removal 4-46
5-1 Conventional Activated Sludge Plant 5-4
5-2 Step Aeration Plant 5-6
5-3 Comparison of Solids Loading on the Final Clarifier for
Conventional and Step Aeration Plants 5-7
5-4 Contact Stabilization Plant 5-8
5-5 Complete Mix Plant 5-12
5-6 Two-Stage Activated Sludge Plant 5-16
IX
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LIST OF FIGURES - Continued
Figure No. Page
5-7 Schematic Diagram of Multi-Stage Oxygenation System 5-18
5-8 Oxygen Consumption for Pure Oxygen Systems Related to
Organic Loading; (F/M) 5-21
5-9 BOD Removal Characteristics for Various Complete Mix
Activated Sludge Plants 5-23
5-10 Relationship Between BOD Removal Rate Constants and
Loading Ratios for the Activated Sludge Modifications 5-25
5-11 Bench-Scale Aeration Unit 5-30
5-12 Schematic of a Continuous-Flow Pilot Unit 5-31
5-13 Determination of Oxygen Uptake Requirements 5-34
5-14 Determination of Sludge Production Characteristics 5-35
5-15 Upgrading a Conventional Activated Sludge Plant to Step
Aeration 5-38
5-16 Upgrading a Conventional Activated Sludge Plant to
Contact Stabilization 5-41
5-17 Upgrading a Contact Stabilization Plant to Complete-Mix
Activated Sludge: 5-44
5-18 Upgrading a Modified Aeration Activated Sludge System to
Oxygen-Activated Sludge 5-49
5-19 Upgrading a Primary Treatment Plant to Pure Oxygen
Activated Sludge Treatment 5-50
6-1 Typical Clarifier Configurations 6-6
6-2 Clarifier Modifications at the Greater Peoria Sanitary District
Sewage Plant 6-9
7-1 Typical Cross Section of a Facultative Lagoon 7-5
7-2 Typical Microscreen Unit 7-8
7-3 Typical Pilot Plant Data for Filter Design 7-18
7-4 Effect of Reactivation on Adsorption Capacity 7-27
7-5 COD Isotherms Using Virgin Carbon and Different Secondary
Wastewater Effluents 7-31
7-6 Total Capital Costs i'or Carbon Treatment 7-33
7-7 Carbon Adsorption Operation and Maintenance Costs 7-34
7-8 Total Annual and Unit Costs for Carbon Treatment 7-35
8-1 Primary Treatment Units at the Central Contra Costa Sanitary
District Water Reclamation Plant 8-3
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LIST OF FIGURES - Continued
Figure No. Page
8-2 Various Postaeration Devices 8-6
8-3 Cascade Aerator at Pittsfield, Mass. 8-11
9-1 Relationship Between Concentration and Time for 99 Percent
Destruction of Escherichia Coli by Different Forms of
Chlorine at 2 to 6°C 9-3
9-2 MPN Coliform Vs. Chlorine Residual 9-4
9-3 Impact of Chlorine Tank Baffle Design on Actual Detention
Time 9-7
10-1 Sludge Treatment Processes 10-2
10-2 Typical Sludge Processing Systems Using Biological Stabilization 10-4
10-3 Typical Sludge Processing Systems Using Nonbiological
Stabilization 10-5
10-4 Upgrading Sludge Handling Facilities Using Gravity Thickening 10-13
10-5 Schematic of an Air Flotation Unit 10-15
10-6 Upgrading Sludge Handling Facilities Using Air Flotation
Thickening 10-18
10-7 Upgrading Sludge Handling Using Centrifugal Thickening 10-24
11-1 Relationship Between pH and Bicarbonate Concentration 11-4
11-2 Anaerobic Digestion Systems 11-7
11-3 Relationships Between Sludge Solids Digester Loadings, and
Detention Time 11-11
11-4 Upgrading an Existing Low-Rate Digestion System Using
Pre-Thickening of the Combined Sludge and Improvements
to the Primary Digester 11-16
11-5 Conversion of Anaerobic to Aerobic Digestion 11-24
12-1 Typical Vacuum Filter Flow Diagram 12-2
12-2 Types of Centrifuges 12-18
13-1 Case History No. 1 Comparison of Original and Upgraded Flow
Diagrams 13-2
13-2 Hydrasieve Screening Unit 13-4
13-3 Case History No. 2 Comparison of Original and Upgraded Flow
Diagrams 13-6
13-4 Case History No. 4 Comparison of Original and Upgraded Flow
Diagrams 13-10
13-5 Case History No. 5 Comparison of Original and Upgraded Flow
Diagrams 13-14
xi
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LIST OF FIGURES - Continued
Figure No. Page
13-6 Case History No. 6 Comparison of Original and Upgraded Flow
Diagrams 13-20
13-7 Case History No. 7 Comparison of Original and Modified Flow
Diagrams 13-27
13-8 Case History No. 8 Flow Diagram 13-29
13-9 Upgrading a Modified Aeration System for Nutrient Removal
Flow Diagram Primary and Secondary Systems 13-33
13-10 Upgrading a Modified Aeration System for Nutrient Removal
Flow Diagram Nitrification and Denitrification Systems 13-34
13-11 Upgrading a Modified Aeration System for Nutrient Removal
Flow Diagram Filtration and Disinfection Systems 13-35
XII
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LIST OF TABLES
Table No. Page
2-1 Characteristics of Septage 2-7
3-1 Effect of Flow Equalization on Primary Settling .
Newark, New York 3-5
3-2 Cost of Equalization Facilities 3-16
4-1 Operating Data for Single-Stage Trickling Filter Plants at Low,
Intermediate, and High Loading Rates 4-4
4-2 Trickling Filter Volumes for Various Organic Removals as
Calculated by Different Design Equations 4-10
4-3 Comparative Physical Properties of Trickling Filter Media 4-15
4-4 Operating Data for Pueblo, Colorado 4-20
4-5 Operating Data for Chapel Hill, North Carolina 4-23
4-6 Operating Data for Kankakee, Illinois 4-25
4-7 Upgrading Techniques for Improvement of Trickling Filter Plant
Efficiency 4-27
4-8 Trickling Filter Nitrification Data 4-29
4-9 Operating and Design Conditions for Allentown, Pennsylvania 4-34
4-10 Operational and Design Data for an Industry in New York State 4-37
4-11 Summary of Treatment Processes for Nutrient Removal 4-38
4-12 Operating Data for Richardson, Texas 4-40
4-13 Operating Data for Chapel Hill, North Carolina 4-45
4-14 Operating Data for Marlborough, Massachusetts 4-47
5-1 Operating Data from Conventional Activated Sludge Plants 5-3
5-2 Operating Data from Step Aeration Activated Sludge Plants 5-5
5-3 Operating Data from Contact Stabilization Activated Sludge Plants 5-9
5-4 Suggested Design Guidelines 5-10
5-5 Operating Data from Complete Mix Activated Sludge Plants 5-13
5-6 Operating Data from Modified Aeration Activated Sludge Plants 5-14
5-7 Operating Data from Two-Stage Activated Sludge Plants 5-17
5-8 Operating Data from Pure Oxygen Activated Sludge Plants 5-20
5-9 Oxygen and Air Requirements for Activated Sludge Modifications 5-26
5-10 Oxygen Transfer Capabilities of Various Aeration Systems 5-27
5-11 Operating and Performance Data for the Wards Island Plant,
New York City 5-39
xui
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LIST OF TABLES - Continued
Table No. Page
5-12 Operating and Performance Data for Austin, Texas 5-42
5-13 Operating and Performance Data for Coralville, Iowa 5.45
5-14 Operating and Performance Data for Newtown Creek Plant,
Brooklyn, New York 5-47
5-15 Operating and Performance Data for Fairfax County, Virginia 5-31
5-16 Upgrading Techniques for Improvement of Activated Sludge
Treatment Plant Efficiency 5-53
5-17 Design Guidelines for Nitrification 5-54
6-1 Typical Design Parameters for Primary Clarifiers 6-3
6-2 Typical Design Parameters for Secondary Clarifiers 6-4
6-3 Tube Settler Installations 6-15
6-4 Effect of Chemical Treatment on Primary Clarifier Performance 6-17
6-5 Polyelectrolyte Addition to Primary Clarifiers 6-19
6-6 Lime Addition to Primary Clarifiers 6-19
6-7 Effect of Chemical Treatment on Secondary Clarifier Performance 6-21
7-1 Operational Data from Shallow Aerobic Polishing Lagoons 7-2
7-2 Removal Efficiencies for Deep Aerated Effluent Polishing Lagoons 7-3
7-3 Mechanical Mixing Energy Required for Oxygen Dispersion 7-4
7-4 Comparison of Operational Data from Facultative and
Aerated Polishing Lagoons 7-6
7-5 Microscreen Fabric Sizes 7-7
7-6 Typical Microscreen Power and Space Requirements 7-9
7-7 Typical Microscreen Design Parameters 7-11
7-8 Microscreen Performance Data 7-13
7-9 Typical Multimedia Gradations 7-14
7-10 Factors Governing Filter Performance 7-15
7-11 Expected Filter Performance for Activated Sludge Plants 7-19
7-12 Expected Filter Performance for Trickling Filter Plants 7-19
7-13 In-Depth Filtration of Activated Sludge and Trickling Filter
Plant Effluents 7-20
7-14 Surface Filtration of Activated Sludge and Trickling Filter
Plant Effluents 7-21
7-15 Filtration of Chemically Treated Secondary Effluent 7-22
7-16 Typical Carbon Column Design Data 7-25
7-17 Performance of Upflow Bed and Downflow Bed Adsorbers 7-28
7-18 Performance of Tertiary Carbon Wastewater Treatment Plants 7-29
xiv
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LIST OF TABLES - Continued
Table No. Page
8-1 Performance of Primary Facilities at Central Contra Costa
Sanitary District with Chemical Treatment to Preaeration 8-5
9-1 Chlorine Dosage Ranges 9-5
9-2 Relative Total Annual Costs of Disinfection Alternatives 9-10
9-3 Effect of Chlorine on Odor Reduction for a Raw Domestic
Wastewater 9-13
9-4 Effect of Powdered Activated Carbon on Odor Reduction 9-14
9-5 Capital Costs for Odor Control Systems 9-15
10-1 Thickener Design Loadings and Underflow Concentrations 10-9
10-2 Example of Upgrading Sludge Handling Facilities Using
Gravity Thickening 10-12
10-3 Typical Air Flotation Design Parameters 10-14
10-4 Air Flotation Thickening Performance Data 10-16
10-5 Example of Upgrading Sludge Handling Facilities Using Air
Flotation Thickening 10-17
10-6 Centrifugal Thickening Performance Data 10-21
10-7 Example of Upgrading Sludge Handling Facilities Using
Centrifugal Thickening 10-22
11-1 Concentrations Which Will Cause a Toxic Situation in
Anaerobic Digestion of Municipal Sludges 11-6
11-2 Typical Design Criteria for Low-Rate and High-Rate Anaerobic
Digesters 11-8
11-3 Typical Properties of Anaerobic Digester Supernatant 11-13
11-4 Upgrading an Existing Low-Rate Digestion System Using
Prethickening of the Combined Sludge and Improvements
to the Primary Digester 11-15
11-5 Typical Properties of Aerobic Digester Supernatant 11-17
11-6 Batch-Type Aerobic Sludge Digestion Operating Data for
Mixtures of Primary and Waste Activated Sludge 11-19
11-7 Results of High-Purity Oxygen Aerobic Digestion Speedway,
Indiana 11-20
11-8 Aerobic Digestion Design Parameters Using Air 11-22
11-9 Aerobic Digester Upgrading Design Parameters 11-25
11-10 Characteristics of Heat Treated Sludge Filtrate at Batavia 11-26
11-11 Bactericidal Effect of Lime Addition to Chemically
Precipitated Sludges 11-27
xv
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LIST OF TABLES - Continued
Table No. Page
12-1 Typical Specific Resistance Values for Various Sludges 12-6
12-2 Vacuum Filtration Performance Using Inorganic Chemicals
and Purifloc C-31 on Municipal Sludge 12-8
12-3 Summary Data on Vacuum Filtration of Pure Oxygen
Aeration Sludges 12-9
12-4 A Comparison Between Lime/Ferric Chloride and Poly electrolytes
for Conditioning Raw Primary Sludge 12-12
12-5 Sludge-Drying Bed Area Requirements 12-15
12-6 Summary of Centrifuge Characteristics 12-17
12-7 Solid Bowl Centrifuge Performance Data 12-22
13-1 Case History No. 1 Plant Operating and Performance Data 13-3
13-2 Case History No. 2 Plant Operating, Performance and
Design Data 13-7
13-3 Case History No. 3 Plant Operating and Performance Data 13-8
13-4 Case History No. 4 Plant Operating and Design Conditions 13-12
13-5 Summary of Treatment Performance for Cast History No. 4 13-13
13-6 Case History No. 5 Plant Operating and Design Conditions 13-15
13-7 Summary of Treatment Performance for Case History No. 5 13-18
13-8 Estimated Costs for Construction for Case History No. 5 13-18
13-9 Case History No. 6 - Plant Design Conditions 13-22
13-10 Summary of Treatment Performance for Cast History No. 6 13-24
13-11 Capital Costs of Upgrading for Case History No. 6 13-25
13-12 Case History No. 7 - Operating Data 13-26
13-13 Case History No. 7 - Performance Data 13-28
13-14 Case History No. 8 Operating and Performance Data 13-30
13-15 Case History No. 9 Anticipated Performance After Upgrading 13-32
xvi
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FOREWORD
The formation of the United States Environmental Protection Agency marked a new era of
environmental awareness in America. This Agency's goals are national in scope and
encompass broad responsibility in the area of air and water pollution, solid wastes,
pesticides, and radiation. A vital part of EPA's national water pollution control effort is the
constant development and dissemination of new technology for wastewater treatment.
It is now clear that only the most effective design and operation of wastewater treatment
facilities, using the latest available techniques, will be adequate to meet the future water
quality objectives and to ensure continued protection of the nation's waters. It is essential
that this new technology be incorporated into the contemporary design of waste treatment
facilities to achieve maximum benefit of our pollution control expenditures.
The purpose of this manual is to provide the engineering community and related industry a
new source of information to be used in the planning, design and operation of present and
future wastewater treatment facilities. It is recognized that there are a number of design
manuals, manuals of standard practice, and design guidelines currently available in the field
that adequately describe and interpret current engineering practices as related to traditional
plant design. It is the intent of this manual to supplement this existing body of knowledge
by describing new treatment methods, and by discussing the application of new techniques
for more effectively removing a broad spectrum of contaminants from wastewater.
Much of the information presented is based on the evaluation and operation of pilot,
demonstration and full-scale plants. The design criteria thus generated represent typical
values. These values should be used as a guide and should be tempered with sound
engineering judgment based on a complete analysis of the specific application.
This manual is one of several available through the Technology Transfer Office of EPA to
describe recent technological advances and new information. This particular manual was
initially issued in October of 1971 and this edition represents the first revision to the basic
text. Future editions will be issued as warranted by advancing state-of-the-art to include new
data as it becomes available, and to revise design criteria as additional full-scale operational
information is generated.
xvu
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CHAPTER 1
INTRODUCTION
The ability of wastewater treatment plants to perform at required levels of efficiency
becomes more critical as water pollution abatement programs achieve their objectives.
Deviations from design performance, which were formerly of lesser consequence, now
become paramount because of their impact on the receiving waters. Improved process
monitoring and plant operation will obviously reduce the incidence of inadequate
performance, but many cases are the result of more basic deficiencies in the treatment
system. Upgrading of wastewater treatment plants may be required to handle higher
hydraulic and organic loadings to meet existing effluent quality and/or to meet higher
treatment requirements. Any of these situations requires optimization of existing facilities
before consideration of additional treatment facilities. It is necessary that a distinction be
made between upgrading to accommodate higher hydraulic and organic loads, and upgrading
to meet stricter treatment requirements. Existing facilities can be made to handle higher
hydraulic and organic loads by process modifications, whereas meeting higher treatment
requirements usually requires significant expansion and/or modification of existing facilities.
Regardless of the cause, the result is that an inadequately treated effluent is discharged. The
historical solution to such a problem has been plant expansion along the same lines as the
original facility, or addition of conventional unit processes to add secondary or tertiary
treatment to the system. Depending on its application, a generalized approach such as this
does not necessarily make optimum use of the previously existing facilities nor of the
expanded facilities. The situation is further complicated where regional treatment systems
are proposed for the future and existing facilities are inadequate for the interim period. In
such cases, a solution must make optimum use of available technology, with minimum
capital expenditure.
Rapid urbanization, development of industries, and stricter treatment requirements often
necessitate unanticipated upgrading of treatment plants or premature implementation of
upgrading programs. Many existing treatment plants are not capable of meeting the more
stringent performance levels required by today's water quality standards. In addition, there
are needs for interim improvements. These considerations, plus economic pressures to
optimize pollution abatement expenditures, make it mandatory that a logical and
technically sound approach to upgrading existing treatment facilities be established. This is
especially true because of the numerous alternatives available for consideration prior to the
selection of a method for upgrading an existing facility. It is for this reason that a major
plant expansion, i.e., complete duplication of existing unit treatment processes, for the
purposes of this manual will be considered the least attractive upgrading procedure available,
since this approach does not consider optimization of existing facilities.
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Therefore, the purpose of this manual will be to present necessary information for
considering various courses of action with regard to an impending or existing plant overload
situation, or with regard to increasing the efficiency to meet stricter treatment standards.
The diversity of causes that necessitate upgrading of existing plants precludes the use of this
manual as a conventional design manual. Therefore, it is aimed at establishing a framework
of possible alternative methods of upgrading overloaded treatment plants. To facilitate the
information presented, only plants treating "typical" domestic wastewaters will be
considered.
Particular upgrading procedures are stressed as interim methods which may be implemented
with a minimum amount of effort and capital expenditure prior to a more elaborate
upgrading or even a major plant expansion. Cost information has been compiled and
estimates prepared for the upgrading of individual unit processes. Capital costs are based on
an EPA cost index of 175, and unless noted otherwise do not include an allowance for land,
right-of-ways, engineering and legal fees, contingencies or interest during construction. When
available, cost information has also been included for the reported case histories on plant
upgrading. Due to the varying complexity of existing plants, the real benefit of the
subsequent cost information will be as a tool for developing comparative capital costs for
various upgrading techniques. Particular unit process cost information must be used
cautiously, since the complexity of the individual situation will dictate the costs required
for upgrading.
The aspects of nutrient removal, although extremely important and oftentimes responsible
for upgrading action at many treatment plants, will be discussed only briefly herein since
more detailed information is presented in other manuals either currently available or soon to
be published.
1-2
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CHAPTER 2
INVESTIGATIVE APPROACH
2.1 Purpose of Upgrading Existing Wastewater Treatment Plants
Upgrading of existing wastewater treatment plants may be required for a number of reasons,
including the following:
1. To meet more stringent treatment requirements
2. To increase hydraulic and/or organic loading capacity
3. To improve poor performance due to improper plant design and/or operation.
The approach to the solution of an upgrading problem will vary depending upon the
objectives to be attained. In some cases, upgrading only a portion of the plant may achieve
significant improvements in plant performance. This chapter describes the general
investigative techniques applicable to most upgrading situations.
2.2 Identification of Existing Problem Areas
As the initial step in solving an upgrading problem, the engineer should thoroughly
familiarize himself with the following aspects of the existing facility:
1. Efficiency of treatment
2. Normal operational and maintenance procedures
3. Condition of structures
4. Condition of process hardware
5. Staffing pattern and level of operator skill.
Thorough discussions with plant operating personnel are of great benefit in determining the
above. These discussions must be supplemented by a complete review of plant operating
records such as daily flow charts, operating logs and laboratory data. Normally, much of this
information is compiled on standard State regulatory agency forms which are submitted
monthly.
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The engineer should become completely familiar with the type of sampling and flow
measurement techniques employed by the plant personnel and verify their accuracy. One
difficult area in assessing plant performance data is the reliability of influent flow
measurement data. Often, plant flow data are obtained from some type of flow recording
instrument, and reliable flow information is possible only when the treatment plant
operator routinely calibrates the flow measuring and recording instruments. A representative
portion of the operating data, normally a period of one to three years, should be evaluated
by the engineer. If adequate operational and performance data are not available, it is the
responsibility of the engineer to collect the additional information needed to complete the
evaluation.
2.3 Upgrading to Meet More Stringent Treatment Requirements
Increased pressure from Federal and State governments, and a more ecologically minded
public are requiring local communities and sanitary districts to enforce existing water
quality standards. In addition, many regulatory agencies are stipulating (1) increased BOD
and SS removal, (2) the maintenance of minimum DO concentrations in the plant effluent
or receiving body of water, (3) more stringent disinfection requirements, (4) phosphorus
removal, (5) partial or complete oxidation of ammonia nitrogen, and, in some cases, (6) high
levels of nitrogen removal.
An essential feature of any upgrading plan is the provision of adequate flexibility in plant
design to accommodate future treatment requirements. These requirements will vary
depending on the location of the treatment plant, the expected useful lifetime of the
existing facility and the condition and projected use of the receiving body of water.
Regardless of the ultimate treatment goal, adequate consideration must be given to
maximum utilization of the existing plant in the overall upgrading plan. In some instances,
moderate improvements in plant efficiency may be sufficient, while in other situations,
extensive improvements may be required to meet very stringent effluent requirements.
Often, upgrading to meet long-term water quality goals involves expansion of major portions
of the plant to accommodate increased flows.
Since this type of upgrading involves a major capital investment, a detailed evaluation of all
feasible alternatives is justified. This evaluation should include a careful analysis of the
increased operating cost and staffing pattern required by the upgraded facility along with
the municipality's ability to provide the needed financial support. Often, major plant
expansion and added tertiary facilities will substantially increase the cost of sludge handling
and disposal, which represents a significant portion of the total plant operating cost.
2.4 Upgrading to Relieve Hydraulic and Organic Overloads
The performance of different unit processes within a treatment plant is affected to varying
degrees by hydraulic and organic overloads. The relationship between higher flows and the
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corresponding changes in BOD and SS loading is also an important consideration. For
example, a significant increase in flow without a corresponding increase in organics will
require a different upgrading approach than when the increase in flow is accompanied by an
increase in BOD and SS.
Hydraulic and organic overloading normally occurs as the tributary population increases
beyond the plant design capacity. Projections must then be made for a reasonable design
period into the future. Forecasting changes in population and wastewater flows in
connection with upgrading of an existing treatment plant may be quite burdensome but
generally will not be subject to as much uncertainty as in similar forecasting for a relatively
undeveloped area. In many cases, the maximum anticipated growth is defined by saturation
of the tributary area. Potential extension of this area must also be considered and is often
limited by topographical constraints and political boundaries. Analysis of local area growth
patterns; examination of local influences such as land use planning studies, zoning
regulations and wastewater discharge ordinances; and full use of State, County and local
planning agencies can all be extremely useful in judging the future expected flows for a
given upgrading situation.
Rapid industrialization in an existing plant's service area can cause increased hydraulic and
organic loads. The alert community, before issuing a building permit, should notify
industries that pretreatment may be required for wastewaters containing toxic or treatment
inhibiting materials, or for wastewaters having an unusually high percentage of organic or
inorganic material compared to typical domestic wastewaters. Equalization of industrial
wastewater discharges may be helpful in minimizing diurnal flow variations and in damping
short-term, high-strength discharges.
Population equivalent equates the organic content or flow contribution of industries to the
ordinary per capita contribution present in domestic wastewaters. Even a relatively small
industry may contribute a significantly higher organic loading than the existing domestic
load. Population equivalents for many industrial wastewaters should be based on COD
analysis rather than BOD, since extremely strong or toxic wastes may show an artificially
low BOD.
Historical data on raw wastewater flows and concentrations can provide valuable
information on the magnitude, frequency and duration of peak flows and loadings.
Electronic data processing and plotting techniques are effective in reducing data to a usable
form and can provide a firm base for projecting future loading conditions (1). Such analyses
can detect whether significant seasonal or weekly variations exist. This is particularly
important in determining whether plant expansions can be designed for annual average
values, or whether weekday industrial loads warrant higher than annual average design values
for sizing of plant components.
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2.4.1 Hydraulic Overloads
Excessive hydraulic loadings in primary and secondary clarifiers are a major cause of
reduced plant efficiency. Increased surface overflow rates and reduced detention periods in
primary clarifiers result in poor BOD and SS capture which increases the loads to be handled
by subsequent treatment units. SS removal in secondary clarifiers is even more sensitive to
excessive hydraulic loadings than primary clarifiers. High secondary clarifier surface
overflow rates in activated sludge plants can result in the direct carry-over of biological
solids from the clarifier sludge blanket into the effluent. In trickling filter plants, high
secondary clarifier loadings prevent effective capture of the biological solids sloughed from
the filter media.
In any upgrading situation, it is essential that a thorough study be made of plant influent
flow records. It is desirable to review actual meter flow charts to determine typical diurnal
flow variations. These data can be used to size flow equalization basins, as discussed in
Chapter 3. Moreover, such analyses are required to determine whether seasonal or
continuous excessive flows exist, and whether the excessive flows are due to groundwater
infiltration or to inflow contributions.
Since excessive flows can hydraulically overload the unit processes in a wastewater
treatment plant, every effort should be made to reduce excessive infiltration/inflow
contributions before undertaking upgrading techniques. Excessive infiltration/inflow is a
likely possibility when influent BOD and SS concentrations are consistently below
150 mg/1. Plant flow and connected population data, and in some cases records of potable
water usage, can be used to determine if an infiltration/inflow problem exists. Major causes
of infiltration are leaky manholes, faulty lateral connections and leaky pipe joints. Major
sources of excessive inflow are illegal downspouts, footer drains, cross-connections with
storm sewers and surface runoff into the top of illogically placed manholes.
Where it has been established that an infiltration/inflow problem exists, detailed study is
recommended (2). Usually this consists of an initial analysis to determine the magnitude and
location of the problem area. A survey is then made to inspect and/or test the suspected
locations. Usually, the affected portions of the sewers are cleaned before the inspection is
begun. Typical testing and inspection techniques include:
1. Smoke testing
2. Hydrostatic testing
3. Air testing
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4. Closed circuit television
5. Photographic methods.
The survey report should identify the specific locations for remedial action, the
recommended rehabilitation method and an estimate of the rehabilitation cost. This cost
can then be compared to the cost of upgrading the plant with and without
infiltration/inflow control to establish a cost effective course of action. The most common
remedial techniques considered are:
1. Internal or external pressure grouting with chemical sealants
2. Manhole grouting
3. Replacing, elevating and/or sealing of manhole covers
4. Replacement of severely damaged sewer sections or service connections
5. Insertion of sewer liners
6. Removal or plugging of illegal inflow connections such as downspouts or footer
drains.
Reported costs for inspection and repair of sewers vary widely due to a number of factors
such as:
1. Sewer age, construction materials and construction methods
2. Accessibility of sewers
3. Sewer size
4. Soil condition.
Because of the great variety of conditions which can be encountered, average cost values or
specific cost data can be misleading. However, a general cost of $5 to $20 per foot has been
estimated (3).
A number of successful infiltration studies and control programs have been reported
(3)(4)(5), and many others are ongoing. Once the problems are corrected, a continuing
program of flow analysis and sewer inspection can significantly prolong the effective
lifetime of the upgraded plant.
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2.4.2 Organic Overloads
The more common causes of organic overloading of existing plants are the addition of
substantial industrial loads, increased discharge of septage to the plant, and in-plant recycle
from poorly operating or overloaded sludge processing operations. Recycled BOD and SS
loads from properly loaded and operated thickeners, digesters, elutriation tanks and vacuum
filters can amount to as much as 25 percent of the total plant organic load. When these
solids processing units become overloaded or are improperly operated, recycled BOD and SS
loads can reach 50 percent or more of the total plant organic load. Unless pretreated or
equalized, or unless the plant was originally sized to account for them, these intermittent
recycle loads can significantly decrease overall plant performance and, when extremely
severe, can cause complete process failure.
Overloads created by industrial contributors can be relieved by pretreatment, by
equalization or by expansion of critical plant components. As discussed earlier, Federal and
State regulations, along with local ordinances, may necessitate pretreatment. Equalization of
industrial flows by the discharging industry, or at the municipal plant, is especially effective
when the discharge is of relatively short duration.
Excessive BOD loadings usually do not have a major adverse effect on primary treatment.
However, in biological treatment systems, BOD overloads can cause anaerobic conditions in
aeration tanks and trickling filters, and may result in unstable, unsettleable activated sludges
which are prone to bulking and odor problems. High influent BOD loads will also result in
excessive secondary sludge production. The engineer evaluating an existing plant should
check all of the normally used loading parameters to determine if additional tank volume,
oxygen supply, and return and waste sludge capacity is necessary, or if operating
adjustments can be implemented.
Excessive SS loadings will result in increased primary and secondary sludge production
which, in turn, may exceed the capacity of existing thickeners, digesters and dewatering
equipment. If adequate plant data exist, the actual loading parameters can be compared to
the plant design values to determine the degree of overloading and to identify alternative
courses of action.
Complete system mass balances for phosphorus and SS are excellent tools to determine the
validity of wastewater and sludge flow measurements and analytical data. Solids balances are
a key factor in identifying the source and magnitude of excessive in-plant recycling loads.
Phosphorus balances are useful because phosphorus is the one major pollutant common to
all municipal wastewaters which does not break down (even partially) to gaseous forms
during biological treatment.
In smaller plants located in communities served partially by septic tank systems, the
discharge of relatively large quantities of septage can create signficant BOD and SS overloads
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and sludge handling and odor problems. Thus, in such an upgrading evaluation, the quantity
of septage which will be discharged to the plant should be thoroughly investigated. In
northern states, where tanks are normally emptied during the spring and summer months,
the annual quantities of materials collected will be introduced to the plant over a 6- to
8-month period. The quantity and characteristics of septage vary widely (5)(6)(7)(8)(9).
This variation is greater in communities that do not employ a permit system, or otherwise
regulate the collection and disposal of septage. In these cases, septage haulers will
indiscriminately include septic tank contents along with raw wastewater collected from pit
toilets, wastes from camping trailer pump-out stations, waste motor oil from service
stations, cutting oil and other hard to treat or toxic wastes from small industries throughout
the community.
Typical concentration ranges for septage are shown in Table 2-1 (7) (8).
TABLE 2-1
CHARACTERISTICS OF SEPTAGE
Parameter Range Average
BOD, mg/1 2,000 - 25,000 9,800
TSS, mg/1 7,000 - 106,000 43,000
VSS, percent of TSS 47 - 82 69
COD, mg/1 5,000 - 80,000 54,000
Two methods are commonly employed to handle septage at existing treatment facilities.
The method of choice will depend on a complete analytical characterization of the septage,
the relative volume of septage to influent wastewater flow and physical constraints at the
plant site.
If the ratio of septage to total plant flow is relatively small on a pound of BOD/day basis,
these wastes may be discharged to an aerated storage basin at the plant site and metered into
the plant at a controlled rate so as not to seriously interefere with normal plant operation
and performance.
If the ratio of septage to total plant flow is high, or if the septage contains toxic or
treatment inhibiting substances, separate treatment facilities must be provided. Full-scale
separate treatment techniques that have been considered include aerobic digestion followed
by sludge dewatering and aerobic digestion followed by introduction of digester contents at
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a controlled rate into the plant. Separate treatment pilot studies have included aerobic
digestion, followed by dewatering on sand beds, polymer thickening, and high lime (pH
11.5) stabilization followed by dewatering on sand beds (7).
2.5 Upgrading to Improve Plant Design and Operation
Many existing plants, although well designed, have not produced the anticipated
performance because of inadequate operator staffing and training, and/or a lack of financial
support and commitment to water pollution abatement from the community. On the other
hand, some poorly designed plants are functioning well because of the excellence of plant
operation. Each upgrading situation will contain its own innate characteristics relative to
plant design and operation. The design engineer must carefully evaluate the
interrelationships of the design and its operation to arrive at the solution which optimizes
both of these factors.
2.5.1 Plant Design
In the past, the problems associated with inadequately designed municipal wastewater
treatment plants have been a major concern of Federal and State agencies. For this reason,
most states have adopted conservative design guidelines and review procedures which must
be followed unless the engineer can substantiate a less conservative viewpoint. The
implementation of these procedures by regulatory agencies has reduced the frequency of
inadequate plant design. However, there are many aspects of detailed design other than
those covered by these regulations which can still adversely affect overall plant performance.
Usually, the plant operators responsible for day-to-day supervision of plant systems and
components can identify major design deficiencies.
Among the numerous design features which can adversely affect plant operation and
performance are:
1. Inadequate standby equipment
2. Fixed speed units where variable speed is essential, such as raw wastewater pumps,
recycle pumps and air blowers
3. Poor hydraulic and solids distribution among identical units operating in parallel
4. Insufficient or inflexible return and sludge wasting pumping capacity
5. Inability of instrumentation and equipment to operate at low flows and loads
occurring in the early lifetime of the plant
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6. Lack of tank dewatering systems to permit rapid servicing of submerged
equipment
7. Lack of flexibility in disinfection systems to permit prechlorination for odor
control, or return sludge chlorination for control of sludge bulking
8. Inadequate electrical and hydraulic capacity to permit standby pumps and air
blowers to operate in parallel
9. Inadequate laboratory facilities
10. Lack of accessibility to key mechanical equipment for routine maintenance or
repair.
From a process standpoint, the most common deficiency is plant design based on average
flow and BOD and SS loadings, with no recognition of peak conditions. In many plants,
sustained peak influent flows and BOD and SS loads can reach two or more times average
values. Therefore, all plant units and systems, including primary and secondary sludge
pumping, air supply equipment and solids processing components, should be designed to
perform successfully at peak conditions. Frequently, peak flows and BOD and SS mass
loadings do not occur simultaneously (10). Thus, a design based on the concurrence of peak
flow and peak pollutant concentration conditions may result in excess capacity. Analysis of
existing plant data, if available, will assist in determining the appropriate peak mass loadings
and flows.
2.5.2 Plant Operation
One of the primary considerations in evaluating an overloaded plant is in the area of plant
operation and control. Therefore, no physical upgrading should be considered before the
engineer is assured that the plant is being operated to yield its maximum efficiency.
A major tool required for proper process control is frequent and accurate sampling and
laboratory analysis. The locations at which samples are taken are critical, particularly in
solids-laden flows where solids tend to travel near the bottom of the conduit.
Many wastewater treatment plants do not have a laboratory equipped to analyze wastewater
samples from the various treatment units in order to assess their performance. Improper
operation, coupled with inadequate laboratory control, increases the probability of poor
treatment. For this reason, plants must be staffed with an adequate number of competent
operators and laboratory personnel. Further, sufficient funds must be made available for
proper maintenance and for the purchase of adequate sampling and analytical equipment
and supplies.
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Because of the availability of field training programs (11) and the trend to mandatory
operator certification, operator skills have improved markedly over the past five years. The
engineer involved in an upgrading situation should identify any areas in which additional
training would result in an improved operation.
2.6 Consideration of Applicable Upgrading Techniques
Technology in the field of wastewater treatment in the past decade has provided many
innovative upgrading procedures. Various research projects sponsored by the U.S. EPA have
resulted in a better understanding of existing and new unit processes. In addition, new types
of equipment have significantly enlarged the range of alternatives available for plant
upgrading.
It has long been recognized that the performance of a wastewater treatment plant is affected
by variations in the influent flow. Equalization of extreme flows can dampen the
fluctuations in flows and mass loadings to existing plants.
Various techniques such as chemical coagulation, polyelectrolyte addition and novel settling
devices are being successfully used to increase removal efficiency in primary and secondary
clarifiers. These procedures can improve removal efficiency while maximizing hydraulic
throughput in the existing facilities.
Several modifications of the conventional activated sludge process, including step aeration,
contact stabilization and complete mix have been successfully used to upgrade various
treatment plants. A most significant recent development in the activated sludge treatment
process is the use of oxygenation as a substitute for air aeration.
Another method of upgrading an overloaded secondary plant is to provide additional
treatment ahead of the existing biological treatment facilities. The use of synthetic media
trickling filters should be considered when roughing treatment would reduce BOD loadings
on existing facilities to acceptable operating levels. Synthetic media filters have been used
successfully as roughing filters in industrial wastewater treatment, and it is very likely that
they will be used in the future for upgrading municipal treatment plants.
The beneficial effect of a nitrified effluent on DO in receiving waters is receiving increasing
attention. For this reason, some regulatory agencies are requiring nitrification of treatment
plant effluents during summer periods, and in some cases are contemplating a year-round
nitrification requirement. Nitrification at higher wastewater temperatures may be
accomplished through modifications to the existing treatment units, such as increasing
aeration tank volume or adding chemicals to the primary clarifier to decrease the organic
loading to existing aeration units. However, dependable year-round nitrification will
normally require a two-stage biological treatment system.
-------
In an increasing number of instances, treatment plants which are functioning satisfactorily
(design flow not exceeded) will have to improve BOD and SS removals because of more
stringent water quality standards. The necessary additional treatment can often be achieved
by polishing the treatment plant effluent. Several methods are currently available and have
been used successfully, including tertiary aerobic and facultative lagoons, microscreening,
multimedia filtration and activated carbon treatment.
Various sludge handling developments which have been successful are: (1) high-rate
anaerobic digestion, (2) aerobic digestion, (3) air flotation thickening or centrifugal
thickening of waste activated sludge, (4) polyelectrolyte addition to improve thickening and
dewatering of sludges and (5) heat treatment processes for sludge stabilization.
Although this manual presents in detail current technology available for upgrading existing
wastewater treatment plants, the engineer should keep in mind the following overall
considerations which will affect the economics of upgrading:
1. The physical condition of existing plant equipment and structures and their
potential uses in an upgrading situation
2. The length of time before a major expansion will be required, based on
population and wastewater flow projections
3. The time required for implementation of various upgrading techniques
4. The compatibility of upgrading procedures with future planned expansion
5. The financial assistance available to the community
6. The costs of the various upgrading techniques that can be used to achieve
essentially the same result. The operation and maintenance costs, as well as the
capital costs, may vary substantially for different available alternatives. Therefore,
economic comparison of these alternatives is essential.
2.7 References
1. Schwinn, D. E., and Dickson, B. H., Computer Analysis of Wastewater Treatment Plant
Operating Data. Metcalf & Eddy, Inc., Internal Document (1970).
2. Sewer System Evaluation. U. S. EPA, Office of Water Program Operations (October,
1973).
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3. American Public Works Association, Control of Infiltration and In-flow into Sewer
Systems. U. S. EPA, Water Pollution Control Research Series, Program No. 11022
EFF, Contract No. 14-12-550, pp. 45-52, p. 93 (December, 1970).
4. Ground Water Infiltration and Internal Sealing of Sanitary Sewers, Montgomery
County, Ohio. U. S. EPA, Water Pollution Control Research Series, Project No. 11020
DHQ (June, 1972).
5. Sewer Bedding and Infiltration - Gulf Coast Area. U. S. EPA, Water Pollution Control
Research Series, Project No. 11022 DEI (May, 1972).
6. Graner, W., Scavenger Waste Disposal Problems on Long Island. Unpublished paper
presented before New York Water Pollution Control Association (June, 1968).
7. Feige, W.A., Oppelt, E.T., and Kreissl, J.F., Alternative to Septage Treatment; Lime
Stabilization Sand Bed Dewatering. Prepublication Interim Report, U. S. EPA,
Advanced Waste Treatment Research Laboratory, National Environmental Research
Center, Cincinnati, Ohio (April, 1974).
8. Kolega, J., Design Curves for Septage. Water and Sewerage Works, 118, pp. 132-135
(May, 1971).
9. Rotondo, V., "Honey Wagon" Sludge Disposal. Journal New England Water Pollution
Control Assoc., 2, No. 1, pp. 59-64 (March, 1968).
10. Schwinn, D. E., and Dickson, B. H., Nitrogen and Phosphorus Variations in Domestic
Wastewater. Journal Water Pollution Control Federation, 44, No. 11, pp. 2059-2065
(November, 1972).
11. Operation of Wastewater Treatment Plants, A Field Study Training Program.
U. S. EPA, Office of Water Programs (1970).
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CHAPTER 3
FLOW EQUALIZATION
3.1 Introduction and Concept
3.1.1 General
The cyclic nature of wastewater flows in terms of volume and strength is well recognized.
Nearly all municipal wastewater treatment plants today are processing variable wastewater
flows. However, improved efficiency, reliability and control is possible when physical,
biological and chemical processes are operated at or near uniform conditions. For this
reason, flow equalization is employed in the field of water supply and in the treatment of
some industrial wastewater. Presently, the advent of more demanding water quality
standards is stirring interest in the application of flow equalization to municipal wastewater
treatment.
The primary objective of flow equalization basins for municipal treatment plants is simply
to dampen the diurnal flow variation, and thus achieve a constant or nearly constant flow
rate through the downstream treatment processes. A desirable secondary objective is to
dampen the concentration and mass flow of wastewater constituents by blending the
wastewater in the equalization basin. This results in a more uniform loading of organics,
nutrients and other suspended and dissolved constituents to subsequent processes.
Through achieving these objectives, flow equalization can significantly improve the
performance of an existing treatment facility, and is a useful upgrading technique. In the
case of new plant design, flow equalization can reduce the required size of downstream
facilities.
3.1.2 Variations of Flow Equalization
Equalization of municipal wastewater flows may be divided into three broad categories:
1. Equalization of dry weather flows
2. Equalization of wet weather flows from separate sanitary sewers
3. Equalization of combined storm and sanitary wastewater.
This chapter is primarily concerned with equalization of dry weather flows. This procedure
provides a technique for achieving normal operation of a treatment plant under near ideal
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loading conditions. Its relatively low cost makes it attractive for upgrading an overloaded
plant.
Increased wet weather flows in sanitary sewers is the sum of two components, infiltration
and inflow. In some cases, it is feasible to equalize stormwater inflow, depending on its
magnitude and duration. Infiltration from high groundwater tables can seldom be equalized.
Equalization of wet weather flows from combined storm and sanitary sewers usually
requires very large storage basins. The design of equalization basins to deal with these types
of flow requires a special knowledge of the collection system, precipitation patterns,
topography, and other factors not directly related to wastewater treatment. Strictly
speaking, wet weather and combined sewer flow equalization cannot be considered as a
wastewater treatment upgrading technique, and the design of such a facility is beyond the
scope of this chapter. However, the concepts presented for dry weather flow equalization
are generally applicable to equalization of wet weather and combined wastewater flows.
In some instances, large interceptor sewers entering the treatment plant can be effectively
used as storage basins to dampen peak diurnal dry weather flow variations. In such cases,
nightly or weekly drawdown of the interceptor system is necessary to flush out solids which
may have been deposited during the previous storage period.
Although the use of influent sewers for equalization should not be ignored, the most
positive and effective means to maximize the benefits possible with equalization is through
the use of specially designed equalization basins. These basins should normally be located
near the head end of the treatment works, preferably downstream of pretreatment facilities
such as bar screens, comminutors, and grit chambers. Adequate aeration and mixing must be
provided to keep the basins aerobic and prevent solids deposition.
It is sometimes desirable to locate the equalization basin at strategic locations within the
collection system. This offers the added advantage of economically relieving trunk sewer
overload during peak flow periods (1). However, it does result in the need for a pumping
facility and therefore is best located where a need for pumping already exists.
Equalization basins may be designed as either in-line or side-line units. In the in-line design
shown on Figure 3-la, all the flow passes through the equalization basin. This results in
significant concentration and mass flow damping. In the side-line design shown on Figure
3-lb, only that amount of flow above the daily average is diverted through the equalization
basin. This scheme minimizes pumping requirements at the expense of less effective
concentration damping.
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FIGURE 3-1
SCHEMATIC FLOW DIAGRAMS OF EQUALIZATION FACILITIES
-CONTROLLED FLOW PUMPING STATION
RAH J
WASTEWATER~*|
BAR SCREEN
AND/OR
COMMINUTOR-
FINAL
'EFFLUENT
FLOW METER AND CONTROL DEVICE
SLUDGE PROCESSING
RECYCLE FLOWS
3-la IN-LINE EQUALIZATION
RAW i | 1 G
IIHSTtllAT tR 1 REM
BAR SCREEN!
AND/OR \
COMMINUTOR-1
CONTROLLED
FLOW PUMPIN
STATION
\
RIT r i
OVAL 1 1
T
EQUALIZA-
TION
BASIN
.H
CONTROL DEVICE^
t
« ^
!/
i i "
i
PRIMARY
TREATMENT
SECONDARY
TREATMENT
FINAL
~~*" EFFLUENT
SLUDGE PROCESSING
RECYCLE FLOWS
3-lb SIDE-LINE EQUALIZATION
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For new construction and for upgrading large plants, it is desirable to construct
compartmentalized or multiple basins. This feature will allow the flexibility to dewater a
portion of the facility for maintenance or equipment repair while still providing some flow
equalization. Where a basin is designed for storage and equalization of wet weather flows,
compartmentalized tanks will allow the utilization of a portion of the basin for dry weather
flow equalization.
Single basin installations may be used for upgrading small plants, but must have the
provision to be dewatered while maintaining complete treatment. This will require a bypass
line around the basin to allow the downstream portion of the plant to operate unequalized
when the flow equalization facility is out of service.
3.2 Benefits of Dry Weather Flow Equalization
Flow equalization has a positive impact on all treatment processes from primary treatment
to advanced waste treatment.
3.2.1 Impact on Primary Settling
The most beneficial impact on primary settling is the reduction of peak overflow rates
resulting in improved performance, and a more uniform primary effluent quality. Flow
equalization permits the sizing of new clarifiers based on equalized flow rates rather than
peak rates. In an existing primary clarifier that is hydraulically overloaded during periods of
peak diurnal flow, equalization can reduce the maximum overflow rate to an acceptable
level. A constant influent feed rate also avoids hydraulic disruptions in the clarifier created
by sudden flow changes, especially those caused by additional wastewater lift pumps
suddenly coming on line.
LaGrega and Keenan (2) investigated the effect of flow equalization at the 1.8 mgd Newark,
New York Wastewater Treatment Plant. An existing aeration tank was temporarily
converted to an equalization basin. They compared the performance of primary settling
under marginal operating conditions, with and without equalization. The results are shown
in Table 3-1.
It has been demonstrated (3) (4) that preaeration can significantly improve primary settling,
as discussed in Chapter 8. Roe (3) concluded that preaeration preflocculates SS thereby
improving their settling characteristics. Indications are that this benefit may be realized by
aerated equalization basins. This benefit may be diminished when the equalized flow is
centrifugally pumped to the primary clarifier, due to the shearing of the floe.
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TABLE 3-1
EFFECT OF FLOW EQUALIZATION ON PRIMARY SETTLING
NEWARK, NEW YORK
Normal Flow Equalized Flow
Primary Influent SS, mg/1 136.7 128.0
Primary Effluent SS, mg/1 105.4 68.0
SS Removal in Primaries, percent 23 47
Note: Average flow slightly higher in unequalized portion of study.
3.2.2 Impact on Biological Treatment
As contrasted to primary treatment or other mainly physical processes where concentration
damping is of minor benefit, biological treatment performance can benefit significantly
from both concentration damping and flow smoothing. Concentration damping can protect
biological processes from upset or failure from shock loadings of toxic or treatment
inhibiting substances. Therefore, in-line equalization basins are preferred to side-line basins
for biological treatment applications.
Improvement in effluent quality due to stabilized mass loading of BOD on biological
systems treating normal domestic wastes has not been adequately demonstrated to date. It is
expected that the effect will be significant where diurnal fluctuations in organic mass
loadings are extreme. This situation may arise at a wastewater treatment plant receiving a
high-strength industrial flow of short duration. Damping of flow and mass loading will also
improve aeration tank performance where aeration equipment is marginal or inadequate in
satisfying peak diurnal loading oxygen demands (5).
The optimum pH for bacterial growth lies between 6.5 and 7.5. In-line flow equalization can
provide an effective means for maintaining a stabilized pH within this range.
Flow smoothing can be expected to improve final settling even more so than primary
settling. In the activated sludge process, flow equalization has the added benefit of
stabilizing the solids loading on the final clarifier. This has two ramifications:
1. The MLSS concentration can be increased thereby decreasing the F/M and
increasing the SRT. This may result in an increased level of nitrification, and a
decrease in biological sludge production. It may also improve the performance of
a system operating at an excessively high daily peak F/M.
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2. Diurnal fluctuations in the sludge blanket level will be reduced. This reduces the
potential for solids being drawn over the weir by the higher velocities in the zone
of the effluent weirs.
3.2.3 Miscellaneous Benefits
In chemical coagulation and precipitation systems using iron or aluminum salts, the quantity
of chemical coagulant required is proportional to the mass of material to be precipitated.
Damping of mass loadings with in-line equalization will improve chemical feed control and
process reliability, and may reduce instrumentation complexity and costs.
Flow smoothing will reduce the surface area required and enhance the performance of
tertiary filters. A constant feed rate will lead to more uniform solids loadings and filtration
cycles.
The equalization basin provides an excellent point of return for recycled concentrated waste
streams such as digester supernatant, sludge dewatering filtrate and polishing filter
backwash.
Some BOD reduction is likely to occur in an aerated equalization basin. A 10 to 20 percent
reduction has been suggested (6) for an in-line basin equalizing raw wastewater. However,
the degree of reduction will depend upon the detention time in the basin, the aeration
provided, wastewater temperature and other factors. For an existing treatment plant, a
simple series of oxygen uptake studies on a representative sample of wastewater can
determine the BOD reduction that will occur.
Roe (3) observed that preaeration may improve the treatability of raw wastewater by
creating a positive oxidation-reduction potential, thereby reducing the degree of oxidation
required in subsequent stages of treatment.
3.3 Determination of Equalization Requirements
The design of an equalization basin requires the evaluation and selection of a number of
features as follows:
1. In-line versus side-line basins
2. Basin volume
3. Degree of compartmentalization
4. Type of construction - earthen, concrete or steel
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5. Aeration and mixing equipment
6. Pumping and control concept
7. Location in treatment system.
The design decisions must be based on the nature and extent of the treatment processes
used, the benefits desired, and local site conditions and constraints.
It may not be necessary to equalize the entire influent flow where high flow or
concentration variations can be attributed to one source, such as an industry. In these cases
the desired benefits can be achieved by simply equalizing the industrial flow. This can be
accomplished through construction of an equalization basin at the industrial site or through
in-house industrial process modifications to effect an equalized wastewater discharge.
3.3.1 Determination of Required Volume
Two methods are available for computing equalization volume requirements. One procedure
is based on the characteristic diurnal flow pattern. In this case, the function of the basin is
to store flows in excess of the average daily flow and to discharge them at times when the
flow is less than the average. The required volume can be determined graphically through
the construction of a hydrograph. The second procedure is based upon the mass loading
pattern of a particular constituent. This method computes the volume required to dampen
mass loading variations to within a preset acceptable range (7) (8).
Since the prime objective of flow equalization in wastewater treatment is to equalize flow,
the determination of equalization volume should be based on the hydrograph. Once the
volume has been determined for flow smoothing, the effect on concentration and mass load
damping can be estimated. The required volumes for side-line and in-line basins will be
identical. The hydrograph procedure is discussed below.
The first step in design involves the establishment of a diurnal flow pattern. Whenever
possible, this should be based upon actual plant data. It is important to note that the diurnal
pattern will vary from day to day, especially from weekday to weekend and also from
month to month. The pattern selected must yield a large enough basin design to effectively
equalize any reasonable dry weather diurnal flow. Figure 3-2 depicts a typical diurnal flow
pattern. The average flow rate is 4.3 mgd. For purposes of this example, the average flow is
used as the desired flow rate out of the equalization basin. The diurnal peak and minimum
flow rate for this example are 1.7 and 0.45 times the average, respectively.
The next step involves the actual construction of the hydrograph. The hydrograph for this
example is shown on Figure 3-3. The inflow mass diagram is plotted first. To do this, the
hourly diurnal flows are converted to equivalent hourly volumes, and accumulated over the
3-7
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FIGURE 3-2
RAW WASTEWATER FLOW AND BOD VARIATION
BEFORE EQUALIZATION
- 300
BOD MASS LOADING
PEAK: AVERAGE =1.97
MINIMUM: AVERAGE = 0.14
PEAK:MINIMUM = 14.59
12
MIDNIGHT
0
12
MIDNIGHT
-_4QO «P
__ 200
TIME OF DAY
3-8
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4,000 -
3,500-
3,000 -
2,500-
2,000-
1 ,500-
,000-
500-
FIGURE 3-3
HYDROGRAPH FOR TYPICAL DIURNAL FLOW
REQUIRED EQUALIZATION
VOLUME, 740,000 GALLONS
INFLOW MASS DIAGRAM
AVERAGE FLOW 4.3 MGD
12 2
MIDNIGHT
NOON
TIME OF DAY
3-9
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24-hour day. A line is then drawn from the origin to the end point on the inflow mass
diagram. The slope of this line actually represents the average flow for the day.
Enough tank volume must be provided to accumulate flows above the equalized flow rate.
This normally requires a volume equivalent to 10 to 20 percent of the average daily dry
weather flow. To determine this volume, the inflow mass diagram must be enveloped with
two lines parallel to the average flow line and tangent to the extremities of the inflow mass
diagram. These are shown as lines A and B on Figure 3-3. The required volume is
represented by the vertical distance between these two lines. In this illustration, the required
volume for equalization is 740,000 gallons, which represents approximately 17 percent of
the average daily flow.
The physical interpretation of the hydrograph is simple. At 8:00 AM, the equalization basin
is empty, as signified by the tangency of the inflow mass diagram with the bottom diagonal.
At this point, plant flow begins to exceed the average flow rate and the tank begins to fill.
This is represented by the divergence of the inflow mass diagram and the bottom diagonal.
At 5:00 PM, the basin is full, as shown by the tangency of the inflow mass diagram with the
top diagonal. Finally, the tank is drawn down from 5:00 PM to 8:00 AM on the following
day, when the flow is below average.
The actual equalization basin volume must be greater than that obtained with the
hydrograph for several reasons, including:
1. Continuous operation of aeration and mixing equipment will not allow complete
drawdown.
2. Volume must be provided to accommodate anticipated concentrated plant recycle
streams.
3. Some contingency should be provided for unforeseen changes in diurnal flow.
The final volume selected should include adequate consideration of the conditions listed
above and will also depend on the basin geometry. For the example presented herein, a
basin volume of approximately one million gallons is adequate.
3.3.2 Impact of Equalization on Diurnal Concentration Variation
At this point, it is appropriate to examine the impact of flow equalization on mass loading
and concentrations. As previously mentioned, side-line equalization has a minimal effect on
diurnal concentration variations. The following discussion is therefore limited to in-line
basins.
3-10
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An hourly concentration plot for raw wastewater BOD is plotted with the diurnal flow
pattern on Figure 3-2. Note that low BOD concentrations occur at night with low flows, and
high BOD concentrations occur during the daytime with high flows. This is a typical pattern
for dry weather flows and BOD's. Because of this characteristic, the mass loading rate of raw
wastewater BOD, shown on Figure 3-2, exhibits even greater fluctuations. If this wastewater
is equalized in a one million gallon in-line basin, the equalized flow will exhibit the
characteristics shown on Figure 3-4, provided:
1. The basin is designed to provide complete mixing.
2. There is no BOD reduction in the basin.
This damping effect would be similarly beneficial for all concentration variables including
SS, nitrogen, phosphorus, and toxic constituents.
On Figure 3-4, the changes in BOD concentration are most pronounced during periods of
minimum wastewater volume in the equalization tank. If desired, increased damping can be
achieved by increasing the active volume of the tank, i.e., the volume in excess of that
obtained from the hydrograph.
3.3.3 Basin Construction
Equalization basins can be provided through the construction of new facilities or through
the modification of existing facilities of sufficient volume. Equalization may be
implemented with relative ease in an upgrading plan that calls for the abandonment of
existing tankage. Facilities which may be suitable for conversion to equalization basins
include aeration tanks, clarifiers, digesters and sludge lagoons.
New basins may be constructed of earth, concrete or steel. Earthen basins are generally the
least expensive. They can normally be constructed with side slope varying between 3:1 and
2:1 horizontal to vertical, depending on the type of lining used. To prevent embankment
failure in areas of high groundwater, drainage facilities should be provided for groundwater
control. In large basins where a combination of aerator action and wind forces may cause
the formation of large waves, precaution should be taken in design to prevent erosion. It is
also customary to provide a concrete pad directly under the equalization basin aerator or
mixer. The top of the dikes should be wide enough to ensure a stable embankment. For
economy of construction, the top width of the dike should be sufficient to accommodate
mechanical compaction equipment.
In-line basins should be designed to achieve complete mixing in order to optimize
concentration damping. Elongated tank design enhances plug flow and should be avoided.
Inlet and outlet configurations should be designed to prevent short-circuiting. Designs which
discharge influent flow as close as possible to the basin mixers are preferred.
3-11
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FIGURE 3-4
RAW WASTEWATER FLOW AND BOD VARIATION
AFTER EQUALIZATION
8 -
7 -
6-
a
3
E
. 5 -
J
C
'- 4 -
C
>
- 3 -
2 -
1 -
n
1
MIDN
BOD MASS LOADING
PEAK: AVERAGE= 1.22
MINIMUM:AVERAGE = 0.61
PEAK:MINIMUM = 2.01
BOD CONC.-^
X\ /' BOD MASS LOADING-v
'\. X\ 1 ^--^ FLOW RATE-/ X
'\. X^^ f'^
2 6 12 6 1
IGHT NOON MIDN
- 30
- 20
- 10
n
2
GHT
-r- 600
-- 400
oo
CO
-- 200
-L-0
TIME OF DAY
3-12
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To continue the previous illustration, an earthen basin has been selected for the equalization
facility. A square plan has been chosen to effect optimum mixing. A section view of the
basin with appropriate dimensions is shown on Figure 3-5. The volume requirement
computed from the hydrograph is provided in the upper eight feet. Note that the minimum
required operating depth lies above the minimum allowable aerator operating level.
3.3.4 Air and Mixing Requirements
The successful operation of both in-line and side-line basins requires proper mixing and
aeration. Mixing equipment should be designed to blend the contents of the tank, and to
prevent deposition of solids in the basin. To minimize mixing requirements, grit removal
facilities should precede equalization basins wherever possible. Aeration is required to
prevent the wastewater from becoming septic. Mixing requirements for blending a municipal
wastewater having a SS concentration of approximately 200 mg/1 range from 0.02 to
0.04 hp/1,000 gallons of storage. To maintain aerobic conditions, air should be supplied at a
rate of 1.25 to 2.0 cfm/1,000 gallons of storage (9).
Mechanical aerators are one method of providing both mixing and aeration. The oxygen
transfer capabilities of mechanical aerators operating in tap water under standard conditions
vary from 3 to 4 Ib 02/hp-hr. Baffling may be necessary to ensure proper mixing,
particularly with a circular tank configuration. Minimum operating levels for floating
aerators generally exceed five feet, and vary with the horsepower and design of the unit.
Low level shutoff controls should be provided to protect the unit. The horsepower
requirements to prevent deposition of solids in the basin may greatly exceed the horsepower
needed for blending and oxygen transfer. In such cases, it may be more economical to install
mixing equipment to keep the solids in suspension and furnish the air requirements through
a diffused air system, or by mounting a surface aerator blade on the mixer.
It should be cautioned that other factors including maximum operating depth and basin
configuration affect the size, type, quantity and placement of the aeration equipment. In all
cases, the manufacturer should be consulted.
3.3.5 Pump and Pump Control Systems
Flow equalization imposes an additional head requirement within the treatment plant. As a
minimum, this head is equal to the sum of the dynamic losses and the normal surface level
variation. Additional head may be required if the basin is to be dewatered to a downstream
location. It may be possible to dewater the basin upstream (e.g., ahead of raw wastewater
pumps) by gravity.
Normally, the head requirement cannot be fulfilled by gravity, thereby requiring pumping
facilities. The pumping may precede or follow equalization. In some cases pumping of both
3-13
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FIGURE 3-5
EARTHEN EQUALIZATION BASIN
153'
-3-0'FREEBOARD
MAX. SURFACE LEVEL
FLOATING AERATOR
EL. 15.0
MIN. REQ'D OPERATING
LEVEL
MIN. ALLOWABLE OPERATING
LEVEL TO PROTECT FLOATING
AERATOR*
CONCRETE SCOUR PAD*
VOLUMES:
EL. 0.0 TO EL. 7.0 APPROXIMATELY 260,000 GAL.
EL. 7.0 TO EL. 15.0 APPROXIMATELY 740,000 GAL,
TOTAL VOLUME =
1,000,000 GAL.
*THESE DIMENSIONS WILL VARY WITH AERATOR DESIGN AND HORSEPOWER
3-14
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raw and equalized flows will be required. Influent pumping will require larger capacity
pumps to satisfy diurnal peaks.
Gravity discharge from equalization will require an automatically controlled flow regulating
device.
A flow measuring device is required downstream of the basin to monitor the equalized flow.
Instrumentation should be provided to control the preselected equalization rate by
automatic adjustment of the basin effluent pumps or flow-regulating device.
3.3.6 Miscellaneous Considerations
The following features are considered to be desirable for the design of the equalization
facility:
1. Equalization should be preceded if possible with screening and grit removal to
prevent grit deposition and rag fouling of equipment in the basin.
2. Surface aerators should be fitted with legs to support and protect the units when
the tank is dewatered.
3. Facilities should be provided to flush solids and grease accumulations from the
basin walls.
4. A high-water level takeoff should be provided for withdrawing floating material
and foam,
5. An emergency overflow should be provided.
3.4 Costs
The development of alternatives for any plant upgrading program should include at least one
which incorporates flow equalization. In all cases, the added cost of flow equalization must
be measured against (1) the savings in cost of modifying downstream processes to accept
diurnal variations and (2) the improved performance that can be achieved by operating
downstream processes under relatively constant loading conditions.
The cost of flow equalization will vary considerably from one application to another,
depending on the basin size, construction selected, mixing and aeration requirements,
availability of land, location of facility, and pumping requirements. Some judgment must be
made on the distribution of pumping costs. Pumping costs for an equalization basin used to
upgrade existing facilities should be charged to the basin.
3-15
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Capital costs for equalization facilities have been estimated by Smith, et al, (10) and are
listed in Table 3-2. The costs for earthen basins include plastic liner and floating mechanical
aerators. The costs for the concrete basins include diffused aeration facilities. Pumping costs
are based on the construction of a separate equalization basin effluent pumping station. The
costs were developed in conjunction with activated sludge treatment system designs and
therefore include a proportional amount of the engineering fees and interest during
construction.
TABLE 3-2
COST OF EQUALIZATION FACILITIES
(EPA INDEX 175)
Earthen Basin Concrete Basin
Plant Basin With Without With Without
Size Size Pumping Pumping Pumping Pumping
mgd mil gal
1
3
10
0.32
0.88
2.40
$124,000
170,000
318,000
$ 72,300
84,000
134,000
$175,000
333,000
779,000
$124,000
247,000
595,000
The construction cost for the earthen equalization basin on Figure 3-5 is estimated at
$80,000. The cost includes excavation, plastic liner, sand subbase, concrete scour pad, dike
fill, underdrain and a 40-hp floating aerator. The costs do not provide for pumping costs,
land costs, engineering and legal fees, nor interest during construction.
3.5 Performance and Case Histories
Little full-scale operating data are currently available to compare the performance of
wastewater treatment plants with and without flow equalization. However, an increasing
number of plant designs are incorporating the use of equalization facilities for upgrading
existing plants and construction of new plants. The following case histories are presented as
examples of equalization basin design.
3.5.1 Ypsilanti Township, Michigan
A flow equalization project at the Ypsilanti Township Sewage Treatment Plant is currently
under way. The treatment facility consists of two adjacent activated sludge plants recently
upgraded from 7.0 mgd to treat a total flow of 9.0 mgd. Two 350,000-gallon digesters have
been converted to equalization tanks. Data will be collected over a two-year study period
for each plant. The flow will be equalized to one plant the first year while background data
3-16
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is collected for the remaining plant. The situation will be reversed the second year, with the
flow being equalized to the second plant while unequalized flow performance data are
collected on the first plant. Comparison of these data will be made to determine the
beneficial effects of flow equalization on each plant.
3.5.2 Fond du Lac, Wisconsin
This case illustrates a situation in which only a portion of the flow is equalized. The City of
Fond du Lac, Wisconsin, presently employs a single-stage trickling filter plant to treat
combined municipal-industrial wastes. Placed in operation in 1950, the plant was designed
to treat an ultimate dry weather flow of 8 mgd and a BOD loading of 12,500 Ib/day. The
facility is presently treating an average of 7.1 mgd with a BOD loading of 24,000 Ib/day,
and hence is organically overloaded. This condition is aggravated by the fact that the waste
discharges from a major industrial contributor (a tannery) are presently concentrated during
daylight hours. The tannery discharges wastes to the treatment plant via a separate force
main. It accounts for about 35 percent of the BOD and 50 percent of the SS into the plant,
and about 15 percent of the influent flow, resulting in a widely fluctuating BOD and SS
diurnal load profile.
The wide fluctuations in organic loading are resulting in reduced performance of the.
trickling filters. This, in conjunction with the advent of more stringent treatment standards,
has rendered this facility inadequate. Plans are presently under way to upgrade the
treatment plant.
This case represents an ideal situation for employing partial equalization in the upgrading
scheme. The volume of the wastes from the tannery is relatively small compared to the total
volume of flow received at the plant, whereas the organic contribution is significant.
Therefore, a relatively small volume equalization tank is all that is required to attain
effective organic load equalization. In addition, because the tannery discharges to the
treatment plant via a separate force main, equalization may be accomplished at the
treatment plant site. The effect of equaKzing the tannery flow over 24 hours is illustrated on
Figure 3-6.
Located at the plant site are six abandoned square anaerobic digesters, each measuring
50 feet by 50 feet by 17.5 feet deep. Four of the units have fixed covers and two have
floating covers. The utilization of these tanks for equalizing the tannery flow was
investigated. The investigations indicate that the four fixed cover tanks would be adequate
for equalization for all but a few days each year when the use of the two additional tanks
would be necessary because of high flows or maintenance.
The conversion of the abandoned digesters to equalization tanks entails complete
modification of the four fixed covered tanks and only minimal modification of the two
tanks which have floating covers. The four fixed covered tanks would each require the
3-17
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\
\
\
-------
installation of a mechanical mixer to maintain solids in suspension, including structural
modifications in order to support the mixers. A ventilation system would be required for
the covered tanks to ensure the safety of plant personnel who may enter the tanks for
purposes of inspection or maintenance. Minor structural repairs and waterproofing of all six
tanks would be necessary to ensure their structural integrity and watertightness. The two
floating covers would be removed and the pipe gallery would be converted to a pump
station.
The cost for converting these units to equalization tanks is estimated at approximately
$440,000. This cost includes process pumping equipment and piping, four mechanical
mixers, tank ventilation system, instrumentation, electrical work, structural renovations and
alterations, and engineering fees.
At present, additional studies are under way to evaluate the feasibility of equalization of
tannery wastes at the tannery in lieu of equalizing these wastes at the plant site.
3.5.3 Walled Lake Novi, Michigan
The Walled Lake-Novi Wastewater Treatment Plant is a new 2.1 mgd facility employing
side-line flow equalization. The treatment plant was placed into operation in 1971. It was
designed to meet stringent effluent quality standards, including (1) a summertime monthly
average BOD2Q of 8 mg/1, (2) a wintertime monthly average BOD2Q of 15 mg/1, and (3)
10 mg/1 of SS. The facility utilizes the activated sludge process followed by multimedia
tertiary filters. Ferrous chloride and lime are added ahead of aeration for phosphorus
removal. Sludge is processed by aerobic digestion, and dewatered on sludge drying beds. A
schematic diagram of this facility is shown on Figure 3-7.
A major factor in the decision to employ flow equalization was the desire to load the
tertiary filters at a constant rate. The equalization facility consists of a 315,000-gallon
concrete tank which is equivalent in volume to 15 percent of the design flow. The tank is
15 feet deep and 60 feet in diameter. Aeration and mixing are provided by a diffused air
system with a capacity of 2 cfm/1,000 gallons of storage. Chlorination is provided for odor
control. A sludge scraper is installed to prevent consolidation of the sludge.
The equalization facility is operated as follows (11). The process pumping rate is preset on
the pump controller to deliver the estimated average flow to the treatment processes. The
flow delivered by these pumps is monitored by a flowmeter which automatically adjusts the
speed of the pumps to maintain the average flow rate. When the raw wastewater flow to the
wet well exceeds the preset average, the wet well level rises, thereby actuating variable speed
equalization pumps which deliver the excess flow to the equalization basin. When the inflow
to the wet well is less than the average, the wet well level falls and an automatic equalization
basin effluent control valve opens. The valve releases enough wastewater to the wet well to
reestablish the average flow rate through the plant. Since this is a new plant as opposed to
3-19
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FIGURE 3-7
WALLED LAKE-NOVI WASTEWATER TREATMENT PiLANT
co
N>
O
FLOW
EQUALIZATION
BASIN
FLUENT
»
1
t
WET
NELL
t
k
v^
4
PROCESS
PUMPS
y^-EQUALIZAT
/^^
^^^ FLOW
/ METERS
/
/ '
i /.
G
ON PUMPS
r- FaCI2
| LIME & POL
t
AERATION
,, ' '
TANKS
fELEt
;TROLYTE
FINAL
CLARIFIERS
/-CHLORINE CONTACT (
/
MULTI-
MEDIA
FILTERS
DMMINUTOR AND AERATED
RIT CHAMBER
, ^
EFFLUENT
FILTER BACKWASH
-------
an upgraded plant, no comparative data exist. However, the treatment facility is typically
producing a highly treated effluent, with BOD and SS less than 4 mg/1 and 5 mg/1,
respectively (9).
3.5.4 Novi Interceptor Retention Basin, Oakland County, Michigan
This case (12) illustrates the utilization of an equalization basin within the wastewater
collection system.
A portion of the wastewater collection system for the City of Novi, Michigan, discharges to
the existing Wayne County Rouge Valley Interceptor System. Due to the existing connected
load on the Wayne County system, Novi's wastewater discharge to the interceptor system is
limited to a maximum flow rate of 4 cfs. This rate was matched by the existing maximum
diurnal flows from the city. In order that additional population could be served, it was
decided to equalize wastewater flows to the interceptor system. By discharging to the
interceptor continuously at an average rate of flow, the total wastewater flows from the
City of Novi to the Wayne County Rouge Valley Interceptor system could be increased by a
factor of 2.6.
An 87,000-cu ft concrete basin was constructed for equalizing flows. The tank has a
diameter of 92 feet and a depth of 10.5 feet. Aeration and mixing are provided by a
diffused air system with a capacity to deliver 2 cfm/1,000 gallons of storage.
A manhole located upstream of the equalization basin intercepts the flow in the existing
Novi trunk sewer. The intercepted wastewater flows into a weir structure which allows a
maximum of 4 cfs to discharge into the Wayne County system. The wastewater in excess of
the preset average overflows into a wet well where it is pumped to the equalization basin.
When flows in the interceptor fall below the preset average, a flow control meter generates a
signal opening an automatic valve on the effluent line of the basin, allowing stored
wastewater to augment the flow.
3.6 References
1. Click, C.N., The Feasibility of Flow Smoothing Stations in Municipal Sewage Systems.
U. S. EPA Project No. 11010 FDI, Contract No. 14-12-935 (August, 1972).
2. LaGrega, M.D., and Keenan, J.D., Effects of Equalizing Sewage Flow. Presented at the
45th Annual Conference of the Water Pollution Control Federation, Atlanta, Ga.
(October, 1972).
3. Roe, F.C., Preaeration and Air Flocculation. Sewage Works Journal, 23, No. 2, pp.
127-140 (1951).
3-21
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4. Seidel, H.F., and Baumann, E.R., Effect of Preaeration on the Primary Treatment of
Sewage. JWPCF, 33, No. 4, pp. 339-355 (1961).
5. Boon, A.G., and Burgess, D.R., Effects of Diurnal Variations in Flow of Settled Sewage
on the Performance of High Rate Activated-Sludge Plants. Water Pollution Control,
pp. 493-522 (1972).
6. Ames, Iowa: Private communication with Dr. E. Robert Baumann, Iowa State
University (December 11, 1973).
7. Bradley, P.R., and Oldshue, J.Y., The Role of Mixing in Equalization. Presented at the
45th Annual Conference of the Water Pollution Control Federation, Atlanta, Ga.
(October, 1972).
8. Wallace, A.T., Analysis of Equalization Basins. Journal of the Sanitary Engineering
Division, ASCE, SA6, pp. 1161-1171 (1968).
9. Smith, J.M., Masse, A.N., and Feige, W.A., Upgrading Existing Wastewater Treatment
Plants, Technology Transfer Design Seminar. Presented at Vanderbilt University
(September 18,1972).
10. Smith, R., Eilers, R.G., and Hall, E.D., Design and Simulation of Equalization Basins.
U.S. Environmental Protection Agency, Internal Publication (February, 1973).
11. Johnson & Anderson, Inc., Operation and Maintenance Manual for Wastewater
Treatment Plant, Walled Lake Arm, Huron-Rouge Sewage Disposal System. Oakland
County D.P.W., Oakland Co., Michigan (June, 1973).
12. Johnson & Anderson, Inc., Operation and Maintenance Manual for Sewage Retention
Reservoir, Novi Trunk Extension No. 1, Huron-Rouge Sewage Disposal System.
Oakland County D.P.W., Oakland Co., Michigan (September, 1973).
3-22
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CHAPTER 4
TECHNIQUES FOR UPGRADING TRICKLING FILTER PLANTS
4.1 General
In 1973 there were more than 3,500 trickling filter plants in the United States serving over
28 million people. In contrast, there were approximately 3,750 activated sludge plants
serving 48 million people (1). In the past, the trickling filter plant has been considered the
ideal plant for populations of 2,500 to 10,000.
Several reasons have justified this popularity. One is its economy, not only in first cost, but
also in operation; another is its relative simplicity of operation, which does not require as
highly skilled operators as activated sludge plants.
Although activated sludge plants provide higher levels of treatment and greater operational
flexibility than trickling filter plants, trickling filter plant performance was considered
adequate where stream assimilative capacity was relatively large in relation to population.
However, increased urbanization and more stringent effluent and water quality standards
will require that many existing trickling filter plants be upgraded.
Upgrading of a trickling filter may be required due to hydraulic or organic overloading,
higher effluent quality requirements, or both. In general, decreasing hydraulic or organic
overloading in existing facilities will not produce a significant increase in BOD and SS
removals above the original design values. However, the effluent polishing techniques
discussed in Chapter 7 may provide enough additional treatment to obtain an effluent of the
desired quality.
4.2 Trickling Filter Processes
Trickling filtration consists of uniform distribution of wastewater over the trickling filter
media by a flow distributor. A large portion of the wastewater applied to the filter passes
rapidly through it, and the remainder trickles slowly over the surface of the biological slime
which forms. BOD removal occurs by biosorption and coagulation from the rapidly moving
portion of the flow and by progressive removal of soluble constituents from the more slowly
moving portion of the flow.
The quantity of biological slime produced is controlled by the available food, and the
growth will increase as the organic load increases until a maximum thickness is reached. This
maximum growth is controlled by hydraulic rate, ventilation, type of media, type of organic
matter, amount of essential nutrients present and the nature of the particular biological
growth.
4-1
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In the past, trickling filters have been classified as either low- (standard), intermediate- or
high-rate filters, depending upon hydraulic and organic loading rates.
4.2.1 L ow- Rate Trickling Filters
Low-rate trickling filters are designed to handle organic loadings of 5 to 20 Ib
BOD/day/1,000 cu ft, and hydraulic loadings of 1 to 4 mgd/acre. In general, low-rate filters
do not use recirculation to maintain a constant hydraulic loading, but use either
suction-level controlled pumps or a dosing siphon. Dosing tanks are small, usually with only
a 2-minute detention time based on twice the average design flow so that intermittent
dosing is minimized. Even so, at small plants, low night-time flows may result in
intermittent dosing. If the interval between dosing is greater than one or two hours, the
efficiency of the process will deteriorate since the character of the biological slime will be
altered due to lack of moisture. Under normal conditions, the BOD removal efficiency of a
low-rate filter and secondary clarifier may average 75 to 85 percent. By the addition of
recirculation during periods of low flows to keep the filter wet, it is possible to increase
overall plant efficiency (2).
Low-rate filters are normally constructed using a 5 to 10 foot depth of stone media. In most
low-rate filters, only the top 2 to 4 feet of the filter media have appreciable biological slime.
As a result, the lower portions of the filter may be populated by autotrophic nitrifying
bacteria which oxidize ammonia nitrogen to nitrite and nitrate forms. If the nitrifying
population is sufficiently well established, and if climatic conditions and wastewater
characteristics are favorable, a well-operated low-rate filter, in addition to providing good
BOD removal, can produce a highly nitrified effluent. The benefits of well-nitrified effluents
for reduction of total oxygen demand in receiving waters are being increasingly utilized in
the formulation of effluent standards.
4.2.2 Intermediate-Rate Trickling Filters
Intermediate-rate trickling filters are generally designed to treat hydraulic loadings of 4 to
10 mgd/acre, including recirculation, and organic loadings, excluding recirculation, range
from 15 to 30 Ib BOD/day/1,000 cu ft. In the past, there have been some cases where the
organic loading in the intermediate range stimulated considerable biological filter growth
and the rate of hydraulic loading was not sufficient to eliminate clogging of the trickling
filter media (2). This clogging situation can be remedied somewhat by utilizing relatively
large stone, 3 to 4 inches in diameter. However, many plants operate in the intermediate
loading range without reported operational problems (2) (3). In practice, many high-rate
filters will operate in the intermediate range during the early low-flow period of their
operating lifetime.
4-2
-------
4.2.3 High-Rate Trickling Filters
High-rate trickling filters have hydraulic loadings of 10 to 30 mgd/acre, including
recirculation, and organic loadings of 30 to 60 Ib BOD/day/1,000 cu ft, excluding
recirculation. Media depths of 3 to 6 feet are commonly employed. In all high-rate filters,
recirculation is used to maintain a relatively constant hydraulic loading. The higher organic
loadings in high-rate filters preclude the development of nitrifying bacteria in the lower
section of the filter. Hence, these plants will seldom exhibit any nitrification, and will
generally not perform as well as low-rate filters.
4.2.4 Super-Rate Trickling Filters
Super-rate trickling filters have evolved as a result of the development of various types of
synthetic media. Past experience has indicated that hydraulic loadings of 150 mgd/acre and
higher, including recirculation, may be accommodated in super-rate trickling filters. The
major application of super-rate filters has been for high-strength wastes and as roughing
units. Synthetic media filters, because of their high surface area per unit volume, can
perform as well as high-rate filters at volumetric BOD loadings of about 50 to 100 lb/1,000
cu ft, and hydraulic loadings of 0.5 to 1.5 gpm/sq ft. A discussion of synthetic media
characteristics is presented in Section 4.4.4.
4.2.5 Trickling Filter Performance Data
Performance data for a number of trickling filters under various operating conditions are
presented in Table 4-1. These data emphasize the wide range of effluent quality that can be
expected depending upon design and operating conditions.
4.3 Trickling Filter Design Considerations
Trickling filters traditionally have been designed using one of several equations developed
over the years from trickling filter plant operating data. The equations incorporate different
combinations of the many variables that most affect trickling filter efficiency.
Unfortunately, no equation has yet been developed which reflects the actual performance of
filters, due to the complex interrelationship of the many variables involved. Since it is
difficult to accurately predict trickling filter performance based on these equations, they
have limited utility in trickling filter plant design. Wherever possible, it is recommended that
treatability or pilot plant studies be used to verify performance predictions based upon the
various design equations.
4.3.1 Trickling Filter Design Equations
The trickling filter design equations which are in general use are those published by the
National Research Council (8), Ten-State Standards (9), Velz (10), Caller and Gotaas (1.1)
4-3
-------
TABLE 4-1
OPERATING DATA FOR SINGLE-STAGE TRICKLING FILTER PLANTS
AT LOW, INTERMEDIATE, AND HIGH LOADING RATES
Plant Location
Low-Rate
Aurora, Illinois
Dayton, Ohio
Durham, N. Carolina
Madison, Wisconsin
Richardson, Texas
Intermediate-Rate
Plainfield, New Jersey
Great Neck, New York
High-Rate
Oklahoma City, Oklahoma
Freemont, Ohio
Storm Lake, Iowa
Richland, Washington
Alisal, California
Chapel Hill, N. Carolina
Influent
Flow
mgd
7.8
34.4
1.8
4.8
1.5
4.3
0.5
16.2
1.8
7.4
2.2
0.6
1.4
Filter
Media
Depth
ft
6
7.5
7
10
6.5
6
4
6
3.3
8
4.5
3.2
4.25
Plant
Influent
BOD
mg/1
117
227
375
248
166
629
187
463
134
690
173
293
166
Recircu-
lation
Ratio
0
0
0
0
0
0.6
1.0
1.0
1.5
2.1
2.8
3.1
2.0
Hydraulic
Loading
mgd/acre
2.1
3.5
1.9
2.4
3.9
2.4
7.8
16.3
19.0
21.5
19.6
20.8
16.3
Organic
Loading
Ib BOD/
day/
l,000cu
4.4
12
13
6.4
13.31
25
20
78
41
62
44
53
19
Final
Effluent
BOD
mg/1
ft
14
33
68
33
20
13
20
66
21
61
20
24
44
Filter and
Final
Clarifier
BOD Removal
percent
80
76
74
76
83
83
83
78
78
84
83
87
43
Reference
4
4
4
4
5
4
6
6
6
6
6
6
7
^Computed based on assumed 30 percent BOD removal in primary clarifiers.
-------
Schulze (12) and Eckenfelder (13) (14). The National Research Council, Velz, Eckenfelder,
and Galler and Gotaas equations are presented and discussed in the following sections.
4.3.1.1 National Research Council (NRC) Equations
The NRC formulation was the result of an extensive analysis of operational records from
stone-media trickling filter plants serving military installations. Based on data analyses, the
NRC recommended the following equations for predicting the performance of stone-media
trickling filters:
First or Single Stage:
v _ 100
1 + 0.0085 VVF )
Second Stage:
E2 = 10Q_
l + 0.0085
I-EI
where:
EI - Percent BOD removal efficiency through the first-stage
filter and clarifier
W = BOD loading (Ib/day) to the first or single-stage
filter, not including recycle
V = Volume of the particular filter stage in acre-feet
F = Recirculation factor for a particular stage, (1 + R)/(l + 0.1R)2
R = Recirculation ratio = recirculated flow/plant
influent flow
E2 - Percent BOD removal efficiency through the second-stage
filter and clarifier
W = BOD loading (Ib/day) to the second-stage
filter, not including recycle
Some of the limitations of the NRC equations are:
1. Military wastewater is characteristically more concentrated than average domestic
wastewaters. Consequently, higher percentage removal efficiencies per unit of
volume were easier to obtain.
4-5
-------
2. The effect of temperature on trickling filter performance is not considered.
3. NRG equations indicate that organic loading has a greater influence on filter
efficiency than hydraulic loading. This is probably because of the concentrated
nature of the wastewaters.
4. Applicability is limited to concentrated domestic wastewaters because no factor is
included to account for differing treatability rates.
5. The equation for second-stage filters is based on the existence of intermediate
clarifiers following the first-stage filters.
A comparative plot of trickling filter operating data with the predicted value using the first
or single-stage NRG equation is shown on Figure 4-1. It is evident from Figure 4-1 that the
use of the NRG equation may result in substantial deviation from the actual performance of
a trickling filter.
4.3.1.2 Velz Equation
In 1948, Velz proposed the first major formulation delineating a fundamental law as
contrasted to previous attempts based on data analysis. The Velz equation relates the BOD
remaining at depth D as follows:
LD
-^ =io-KD
L
where:
L = Total removable BOD, mg/1
Lj) = Removable BOD at depth D, mg/1
D = Filter depth, feet
K = Constant
Removable BOD in the Velz equation is defined as the maximum fraction of applied BOD
removed at a specific hydraulic loading range. When recirculation is used, the total applied
BOD, La after dilution by recirculation, R, may be determined as follows:
La =
Li + RLe
1 + R
where:
= Total influent BOD, not including recirculation, mg/1
= Total effluent BOD, mg/1
4-6
-------
FIGURE 4-1
COMPARISON OF TRICKLING FILTER OPERATING DATA WITH NRC EQUATION
100
80
60
40
20
LEGEND
BETHLEHEM, PA (15)
BURGESS, El AL (16)
DEEDS & DATA (17)
DHOMACK (18)
A McCABE & ECKENFELDER (19)
AGALLER & GOTAAS (in
o NATIONAL RESEARCH COUNCIL (7)
2000
4000
LB BOD/DAY
EQUIVALENT ACRE-FT
6000
(A)
8000
4-7
-------
4.3.1.3 Eckenfelder Equation
Eckenfelder expanded the Velz equation to include other factors. Schulze (12) had
postulated that the time of liquid contact with the biological slime is directly proportional
to the filter depth and inversely proportional to the hydraulic loading rate. Combining the
time of contact with Velz's first order equation for BOD removal and including the effect of
changes in filter depth on the BOD removal per unit of depth, Eckenfelder proposed the
following equation:
Le = *
0 K n 0-67
La 1 + 2-5 D
Qx0.5
La = Influent BOD (including recirculation), mg/1
Le = BOD of unsettled filter effluent, mg/1
With A in acres, D in feet and Q in mgd
4.3.1.4 Galler and Gotaas Equation
In 1964, the last major effort to forecast the performance of stone filters was attempted by
Galler and Gotaas (11) using multiple regression analysis of data from existing plants.
Based on regression analysis, the following equation was developed:
Le = K(iLj + rLe)l-19
(i + r) 0.78 (1 + D) 0.67 a 0.25
where:
0.464 ( 43-560 \
K '
0.28 T0.15
Le = Unsettled filter effluent BOD, mg/1
Lj = Filter influent BOD, mg/1
D - Filter depth, feet
i = Influent flow, mgd
r = Recirculation flow, mgd
a ~ Filter radius, feet.
T = Wastewater temperature, deg C
4-8
-------
The Caller and Gotaas equation recognized the effects of recirculation, hydraulic loading,
filter depth, and wastewater temperature as being important in understanding the
performance of a trickling filter. They further indicated that recirculation improves the
performance of a filter, but established a 4:1 ratio as a practical upper limit for
recirculation.
4.3.2 Applicability of Various Trickling Filter Design Equations
The design engineer has available several equations for trickling filter designs, and the
decision to use one in preference to another is often difficult. The availability of several
equations often raises doubts concerning their validity.
An attempt has been made by Hanumanulu (20) to compare the actual performance of a
12-foot deep stone media trickling filter with that predicted using NRG, Ten-States
Standards, Velz, Eckenfelder, and Galler and Gotaas equations. The filter was operated at a
constant flow without recycle as well as with a 1:1 recirculation ratio. It was found that the
Velz, Ten-States Standards and the NRG equations predicted filter efficiencies that are
closer to observed values when operated without recycle, while the Eckenfelder and Galler
and Gotaas equations predicted efficiencies closer to observed values for filters operated
with recirculation. However, because the filter studied was about double the depth of most
filters, the conclusions may have differed from those that would have been observed with a
shallow filter.
Ordon (21) calculated the volume of filter media required to achieve specified BOD
removals using the NRG, Eckenfelder, and Galler and Gotaas equations. The wastewater
flow, BOD, and temperature were assumed as 1 mgd, 100 mg/1, and 20 deg C, respectively.
The trickling filter volume calculated by the different equations is shown in Table 4-2.
Inspection of Table 4-2 indicates characteristic trends which the designer should be aware of
before using any of these equations. When recirculation was zero, the filter volumes
calculated from the NRG and Eckenfelder equations were essentially the same, while the
Galler and Gotaas equation gave volumes which were significantly different. However, when
recirculation was considered, the NRG design volumes were generally quite conservative,
while the volumes calculated by Eckenfelder and Galler and Gotaas equations were more
nearly the same. In general, the NRG equations would seem to apply when recirculation is
not considered, when seasonal temperature differentials are minor, and when the wastewater
load is highly variable and of high strength.
4.3.3 Laboratory and Pilot-Plant Treatability Studies
As previously indicated, treatabiHty or pilot-plant studies are advantageous in verifying the
performance predicted by design equations. The use of treatability studies for design of
trickling filters has been limited due to the lack of suitable laboratory-scale testing methods,
4-9
-------
TABLE 4-2
TRICKLING FILTER VOLUMES FOR
VARIOUS ORGANIC REMOVALS AS CALCULATED
BY DIFFERENT DESIGN EQUATIONS
(ALL VOLUMES IN THOUSANDS OF CUBIC FEET)
Recirculation 50% BOD Removal
60% BOD Removal
70% BOD Removal
75% BOD Removal
80% BOD Removal
90% BOD Removal
Ratio
0
1
2
3
4
5
6
NRC1
2.7
1.7
1.4
1.2
1.0
0.9
_
ECK2
3.8
0.96
0.42
0.24
0.15
0.12
0.08
G&G3
0.2
0.12
0.12
0.12
0.12
0.12
0.12
NRC
6
3.6
2.8
2.6
2.4
2.2
_
ECK
8.5
2.2
0.95
0.55
0.35
0.24
0.17
G&G
1.2
0.28
0.26
0.26
0.26
0.26
0.26
NRC
15
8.8
6.8
6.1
5.8
5.7
_
ECK
20
5
2.3
1.3
0.8
0.6
0.4
G&G
10
1.8
1.2
0.9
0.9
0.9
0-9
NRC
23
15
11
9.9
9.3
8.8
ECK
32
8
3.5
2
1.5
0.92
0.67
G&G
42
5
2.4
1.8
1.6
1.5
1.5
NRC
40
25
20
18
17
16
_
ECK
58
15
7
4
2.5
1.8
1.4
G&G
300
23
7.3
4.3
3.6
3.0
NRC
210
130
105
90
85
80
_
ECK
290
75
35
20
14
9
6
G&G
400
170
80
45
- National Research Council.
2ECK - Eckenfelder.
3G&G - Caller and Gotaas.
Design Conditions:
Filter Influent Flow I mgd
Filter Influent BOD 100 mg/1
Wastewater Temperature 20 deg C
-------
and has generally been restricted to synthetic media. Pilot units can be rented from the
media manufacturers but will require considerable manpower and funds to obtain the
meaningful data needed for design purposes. Treatability studies for evaluation of
stone-media filter design parameters have not been commonly performed.
However, advances are being made in the development of a practical laboratory-scale
piloting facility for both stone and plastic media. Based on the concept of contact time as
introduced by Schulze, the trickling filter process may be modeled by using an inclined
plane to support biological growth (22).
Wastewater is introduced at a variable rate to the top of the slimed inclined plane. The
plane's inclination may be varied to change the contact time. As previously discussed,
Schulze's equation relates the contact time to the depth and hydraulic loading, as well as to
the physical characteristics of the filter media. BOD removal is then assumed to vary with
the following first-order removal equation:
L^ = e-Kt
Li
where:
Le = BOD of unsettled filter effluent, mg/1
LJ = BOD of filter influent, mg/1
K = Treatability constant
t = Contact time, minutes
The inclined plane method furnishes data on BOD removal, contact time, hydraulic loading
and recirculation ratios.
Since the basic purpose of either a laboratory or pilot-plant evaluation is to study variables
that affect filter performance, any treatability study should be of sufficient duration to
adequately describe the following variables as they affect the filter performance:
1. Applied BOD loading
2. Hydraulic loading
3. Recirculation
4. Wastewater temperature.
4-11
-------
The data thus obtained from treatability studies can be evaluated using the various trickling
filter equations previously discussed.
4.4 Trickling Filter Upgrading Considerations
The performance of a trickling filter plant is determined by the complex interrelationship of
numerous variables. Thus, in any upgrading situation, the acquisition and analysis of
year-round performance data are essential in determining the additional treatment facilities
needed to meet more stringent effluent requirements. In addition to those variables that
directly affect performance of the trickling filter itself, the performance of the plant as a
whole also depends on proper integration of the filter with the other unit processes being
utilized. The more important variables to be considered prior to upgrading any trickling
filter plant are discussed in the following sections.
4.4.1 Wastewater Characteristics
Municipal wastewaters vary in composition and strength, depending on the relative amounts
of industrial wastewater and infiltration present. The rate of BOD removal from a domestic
wastewater in a trickling filter generally exceeds the BOD removal rate from an industrial
wastewater which has a high percentage of dissolved BOD. This is due to the high percentage
of colloids in domestic wastewater, and to the increased ability of the filter to remove this
colloidal material. A reasonable explanation for this is that some of these materials are
removed by biological flocculation and not by oxidation and synthesis of new cells.
A trickling filter plant that requires upgrading due to industrial loads in excess of those
anticipated in the design may be improved by equalization of the industrial flows. Often this
is most economically accomplished at the industrial source and is particularly advantageous
when the entire industrial load is discharged in an 8- or 16-hour period. Under other
conditions, equalization facilities at the treatment plant may be provided as discussed in
Chapter 3 to reduce peak loadings on the trickling filter.
4.4.2 Recirculation
The practice of effluent recirculation can be used to improve the efficiency and operation of
some stone-media trickling filters. For example, it can minimize the operational problems
associated with intermittent dosing of low-rate trickling filters. Recirculation ratios of 0.5 to
4.0 have been used in high-rate filters; Caller and Gotaas (11) have demonstrated that a
recirculation ratio of greater than 4.0 does not materially increase plant efficiency and is
also uneconomical. Normal design practice is to use ratios of 1.0 to 2.0.
Recirculation as applied to synthetic media involves a somewhat different concept than is
applied to stone filters. Various types of synthetic filter media have higher minimum
wetting rates, i.e., a rate of flow per unit area which will induce a biological slime
4-12
-------
throughout the depth of the media. This minimum wetting rate typically ranges from 0.5 to
1.0 gpm/sq ft (30 to 60 mgd/acre), depending on the geometric configuration of the media.
Therefore, recirculation in synthetic media filters is practiced to maintain the desired
wetting rate for a particular medium. Generally, increasing the hydraulic loading
substantially above the minimum wetting rate decreases the BOD removal through the filter
(23).
There are many possible flow configurations which may be used with a single- or two-stage
high-rate trickling filter plant. Some of the more common flow diagrams which have been
presented in the Water Pollution Control Federation MOP No. 8, Sewage Treatment Plant
Design (3) are shown on Figure 4-2. Decisions regarding the use of any one of these flow
configurations in a plant upgrading must be based on its suitability to the existing plant
facilities and an examination of the relative economics.
The hydraulic capacity of the trickling filter distribution and underdrain systems also
requires investigation where changes to the recirculation pattern are proposed. Significant
increases in recirculated flow may exceed the capacity of the distribution mechanism or
overload the filter media or underdrain conduits.
4.4.3 Clarifier Capacity
When modifying an existing trickling filter plant, the effect of increased recirculation on
both the primary and final clarifiers must be considered. Recycle schemes requiring the
recycled flow to pass entirely through the primary and/or secondary clarifiers exert
significantly higher load on these units than schemes where the recycle passes through the
filter alone. Since studies have shown that direct recirculation of trickling filter effluent is as
effective as recycling clarified effluent (24), this is the preferred flow pattern where clarifier
loading is otherwise excessive. Recommended loadings for trickling filter clarifiers are
presented in Chapter 6.
4.4.4 Trickling Filter Media
The physical properties of various types of trickling filter media are shown in Table 4-3.
The properties shown in Table 4-3 which are of greatest interest are specific surface area and
percent void space. Greater surface area permits a larger mass of biological slimes per unit
volume, while increased void space allows for higher hydraulic loadings and enhanced
oxygen transfer. The ability of synthetic media to handle higher hydraulic and organic
loadings is directly attributed to the higher specific surface area and void space of these
media compared to stone media and blast furnace slag.
4-13
-------
FIGURE 4-2
COMMON FLOW DIAGRAMS FOR SINGLE AND TWO-STAGE
HIGH-RATE TRICKLING FILTERS (2)
SINGLE-STAGE
1
'
s
mums
1
±_
J
1
;$$^^-
^^
\
_*
i
SLUOSE «ET»««
RECIRCULATED FLOW
PRIMARY CLARIFIER
NOTE 'RCPRINIED IITH PERIISSION FROII ' SEIAGE TREATAEIT PLIBT DESICN
HAHUAL OF PRADTIVE NO B IA7ER POLL, COKIBOL FE«E«»TH«
IASHIICTOI, 0 C IAKIIAL OF ENC PDACTICE »0 36. AMER SOC
CIVIL ENGR , NEW YORK H Y (1950)
CZ3
/ N TRICMING FILTER
IHTEIIEDIATE CLIRIFIER
FINAL CLARIFIER
4-14
-------
TABLE 4-3
COMPARATIVE PHYSICAL PROPERTIES OF TRICKLING FILTER MEDIA
Media
Plastic
Redwood
Granite
Granite
Blast Furnace Slag
Nominal
Size
in.
24 x 24 x 48
47% x 47'/2 x 35%
1-3
4
2-3
Units
per
cuft
2-3
50-60
Unit
Weight
Ib/cu ft
2-6
10.3
90
68
Specific
Surface
Area
sq ft/cu ft
25-35
14
19
13
20
Void Space
percent
94-97
76
46
60
49
An important consideration in any upgrading involving increased recirculation is the effect
of the resulting increased flow through the filter media. In some cases, replacement of an
existing filter media with a media having greater void space may be required to avoid
ponding due to the higher hydraulic loadings and increased biological growths.
4.4.5 Trickling Filter Depth
Low-rate trickling filters have traditionally been designed with depths ranging from 5 to
10 feet. Because the rate of biological activity is greatest at the surface of a stone-media
trickling filter and diminishes with depth, high-rate stone media filters are designed with
shallower depths of 3 to 6 feet to maximize the rates of treatment per unit volume of filter
media.
As previously mentioned, nitrifying bacteria residing in the lower portions of the relatively
deep low-rate filters can produce a highly nitrified effluent, while high-rate filters seldom
exhibit significant nitrification because of high volumetric BOD loadings. The high loadings
favor the growth of heterotrophic carbonaceous organisms rather than autotrophic
nitrifying organisms.
Recommended depths for synthetic media are 15 to 30 feet. This is because optimum
synthetic media hydraulic wetting rates permit smaller diameter units to be used. Thus,
increased depth is required to provide adequate contact time with the wastewater. The low
unit weight of synthetic media enhances the construction of much deeper filters than is
feasible with stone media.
4-15
-------
4.4.6 Ventilation
Proper ventilation of trickling filters is essential to the maintenance of aerobic conditions
throughout the filter media. It is recommended that all drains, channels, and pipes be sized
such that not more than 50 percent of their cross-sectional area will be submerged at the
peak hydraulic loading.
If the trickling filter is constructed on or near grade, provision for ventilation will be less
critical than if the topography necessitates construction well below grade. In these latter
instances, forced ventilation or ventilation shafts may be necessary.
4.4.7 Temperature of Applied Wastewater
The efficiency of trickling filters is affected by wastewater temperature changes in
accordance with the following relationship (25) (26):
ET = E20 e T-20
where:
9 - Constant varying from 1.035 to 1.041
ET = Filter efficiency at temperature, T
E20 = Filter efficiency at 20 deg C
T = Wastewater temperature, deg C
In northern regions, the effect of both air and wastewater temperatures on trickling filter
performance is very evident, and significant deterioration of the plant performance can be
expected during the winter months (27). The effect of air temperature is especially
pronounced in high-rate filters due to the cooling effect of recirculation, and therefore
should be taken into consideration wherever an upgrading involves significant changes in the
recirculation pattern. It has been reported that covering of filters in cold climates does not
substantially increase the performance because the filter covering does not increase the
temperature of the applied wastewater (28).
4.4.8 Sludge Handling
Upgrading secondary treatment facilities usually results in an increase in sludge production.
Prior to any trickling filter upgrading, therefore, a thorough evaluation of the effects of the
upgrading on the sludge processing system should be made and the results incorporated into
the overall plant upgrading plan. Sludge production increases may be particularly significant
where chemical precipitation in the secondary clarifiers is practiced.
4-16
-------
Humus sludge from trickling filters is commonly returned to the primary clarifier, often as
part of the recirculation pattern. Conservative primary clarifier overflow rates must be used
in such cases. Where primary tank overflow rates are excessive, separate gravity thickening
of combined primary and humus sludge can be employed.
Although sludge processing modifications and/or expansions may be costly, an efficient
sludge handling system is crucial to good treatment plant operation. Upgrading
considerations for sludge processing facilities are discussed in detail in Chapters 10 through
12.
4.5 Trickling Filter Upgrading Techniques and Design Basis
Upgrading to relieve overloaded conditions, to improve organic removal efficiency, to
provide nitrification, and to remove nutrients is covered in the following sections. The
choice between the available options will depend on such factors as plant hydraulics,
conditions of existing treatment facilities and future effluent requirements. The final
selection should maximize the utilization of available clarifiers and minimize modifications
in process piping and additional pumping.
4.5.1 Upgrading to Relieve Organic and Hydraulic Overloading
Trickling filter plants may be upgraded to relieve hydraulic and/or organic overloading by
any one of the following three general procedures:
1. Upgrading existing single-stage filters by the construction of additional trickling
filters in parallel with existing units, or by conversion of a low-rate filter to high
rate
2. Upgrading single-stage trickling filters to a two-stage biological system by the
addition of second-stage trickling filters or an activated sludge system
3. Upgrading existing two-stage trickling filters to a multiple-stage biological system.
Upgrading of single-stage filters by the construction of parallel single-stage units or by
conversion to high-rate filters is a straightforward procedure and is not discussed in detail
since all of the factors involved have been covered previously. It is emphasized that before a
decision is made to adopt this procedure, a careful analysis of plant performance data
should be made to determine if effluent criteria can be met.
4-17
-------
4.5.1.1 Upgrading a Single-Stage Low-Rate Trickling Filter by Improving
Distribution
Upgrading an organically overloaded and hydraulically underloaded single-stage low-rate
trickling filter may be accomplished by providing recirculation. This upgrading procedure
was used to improve the wastewater treatment plant performance at Pueblo, Colorado (29).
The original secondary treatment facilities consisted of low-rate rock-media trickling filter
and final clarifier units as shown on Figure 4-3. At the time of design it was determined that
a 3-foot filter depth, without recirculation, would provide adequate treatment to meet
Colorado State Health Department regulations prohibiting a BOD discharge in excess of
30 mg/1. Table 4-4 shows operating data for the plant. It is apparent from these data that
the trickling filters were not operating as well as had been expected. Hydraulic loading was
inadequate to keep the distributor arms moving during periods of low flow. As a result,
sufficient filter growth could not be maintained throughout the media to adequately treat
the relatively high organic loads, especially during the cold weather months when reduced
biological activity and freezing occurred. The shallow depth of the filter media aggravated
these problems.
The plant was upgraded in 1967 to include recirculation facilities having a capacity of about
30 percent of the average daily flow. A flow diagram of the upgraded plant is also shown on
Figure 4-3. Recirculation during periods of low flow provided the necessary flow to keep
the distributors revolving and thus prevented loss of the filter growth. The benefits of
recirculation were particularly evident in the summer months when the process was not
adversely affected by low wastewater temperature. Operating data for the upgraded plant
are shown in Table 4-4. Despite an increase of more than 80 percent in filter organic
loading, effluent BOD was slightly improved.
The construction costs for this upgrading were estimated to be $129,000 and were allocated
as follows:
Recirculation pumps, motors and controls $ 61,000
Piping and valves 68,000
Total $129,000
4.5.1.2 Upgrading a Single-Stage Trickling Filter to a Two-Stage Filter System
Upgrading an organically overloaded single-stage trickling filter may be accomplished by
conversion to a two-stage filtration system. The objective of a pilot-plant study conducted
at the Chapel Hill, N. C., treatment plant in May-July, 1972, was to determine the degree of
improvement that could be obtained by such a conversion (30).
4-18
-------
FIGURE 4-3
UPGRADING A SINGLE-STAGE LOW-RATE TRICKLING FILTER
BY IMPROVING DISTRIBUTION
PRIMARY EFFLUENT
14.6 MGD
SECONDARY
CLARIFIER
EFFLUENT
SLUDGE
TREATMENT SYSTEM BEFORE UPGRADING
PRIMARY EFFLUENT
12,8 MGD IT*
EXISTING
SECONDARY
CLARIFIER
SLUDGE
RECIRCULATION 3.75 MGD
TREATMENT SYSTEM AFTER UPGRADING
NEW RECIRCULATION
PUMPING STATION
EFFLUENT
4-19
-------
TABLE 4-4
OPERATING DATA FOR PUEBLO, COLORADO
Average
Operating
Condition
(1959-66)
Description
Flow, mgd
Influent BOD, mg/1
Influent SS, mgA
Primary Clarifier
Overflow Rate, gpd/sq ft
BOD Removal, percent
SS Removal, percent
Trickling Filter
Depth, ft
Hydraulic Loading, mgd/acrel
Organic Loading, Ib BOD/day/1,000 cu ft1
Recirculation Ratio
Secondary Clarifier
Overflow Rate, gpd/sq ftl
BOD Removal, percent
SS Removal, percent
Overall Plant Performance
BOD Removal, percent
SS Removal, percent
Effluent BOD, mg/1
Effluent SS, mg/1
1 Includes recirculation.
The single- versus two-stage studies were conducted using two pilot rock-media trickling
filter systems operated as shown on Figure 4-4. During the tests, the influent flow to the
single-stage unit was maintained at 1.2 gpm while the two-stage influent flow rate was set at
2.4 gpm. These rates of flow were established so that performance could be compared for
both a single-stage system where the plant influent is split into equal portions for treatment
through the filters in parallel, and a two-stage system where the entire plant flow passes
through the two filters in series. In the two-stage sequence, the entire influent flow was
14.6
148
163
400
47
49
3
9.4
42
0
775
50
70
74
82
39
29
Upgraded
Operating
Condition
(1972)
12.8
273
322
350
50
49
10.7
77
0.3
880
73
75
86
88
37
40
4-20
-------
treated through one primary tank, then through one filter with recirculation to the head of
primary. Effluent from the first-stage filter was then passed through the other primary
clarifier, now serving as an intermediate clarifier. The partially treated flow was then fed to
the second-stage filter with recirculation directly around the filter. Lastly, second stage filter
effluent was split evenly between the two final clarifiers. The operating data given in
Table 4-5 show that 87 percent of the BOD and 93 percent of the SS were removed by the
two-stage system. These were better reductions than the 80 percent BOD and 85 percent SS
removals obtained with single-stage treatment.
4.5.1.3 Upgrading a Single-Stage Trickling Filter to a Two-Stage Biological
Filtration/Activated Sludge System
If the hydraulic and organic loads to a high-rate trickling filter unit are such that it does not
produce the desired effluent BOD quality, it is possible to upgrade the facility by the
addition of an activated sludge unit immediately downstream from the existing filters. In
this situation, the existing trickling filter acts as a roughing filter, and the subsequent
activated sludge unit provides the treatment capacity needed to obtain the desired BOD
removal.
Alterations made on the Kankakee, Illinois, treatment plant beginning in 1968 exemplify
this type of upgrading (31). The flow diagram of the plant before upgrading appears on
Figure 4-5, and operating data for the year preceding the upgrading are summarized in
Table 4-6. The plant was upgraded by conversion of the existing trickling filters to roughing
filters, addition of a conventional activated sludge system with expanded sludge processing
facilities, and modification as necessary to the existing piping. The flow diagram for the
upgraded plant appears on Figure 4-5, and operating data for a year after the upgrading are
given in Table 4-6. Implementation of this upgrading procedure improved the overall plant
performance by increasing the BOD removal from 67 to 98 percent and the SS removal
from 73 to 99 percent. Details concerning the design of an activated sludge system treating
an effluent from a single-stage biological treatment process are presented in Chapter 5.
The construction costs for this plant modification were estimated to be $3,468,000 and
were allocated as follows:
Materials and earthwork $1,905,000
Piping and mechanical 384,000
Equipment 922,000
Electrical 257,000
Total $3,468,000
4-21
-------
FIGURE 4-4
MODIFYING A SINGLE-STAGE TRICKLING FILTER
TO A TWO-STAGE FILTRATION SYSTEM
INFLUENT
1 2 GPM
RECIRCULATION 2 4 GPM
PRIMARY
CLARIFIER
TRICKLING
FILTER
RECIRCULATION PUMP
EFFLUENT
FINAL
CLARIFIER
TREATMENT SYSTEM PERFORMING AS
SINGLE-STAGE TRICKLING FILTER SYSTEM
(ONE OF TWO DUPLICATE TRAINS)
1ST STAGE RECIRCULATION 2.4 GPM
INFLUENT
2.4 GPM
RECIRCULATION PUMPS-
2ND STAGE RECIRCULATION 2,4 GPM
PRIMARY
CLARIFIER
EFFLUENT
1ST STAGE INTERMEDIATE 2ND STAGE FINAL
TRICKLING CLARIFIERS
FILTER (TWO)
TRICKLING SETTLING
FILTER (CONVERTED
PRIMARY
CLARIFIER)
TREATMENT SYSTEM MODIFIED TO
TWO-STAGE TRICKLING FILTER SYSTEM
4-22
-------
TABLE 4-5
OPERATING DATA FOR CHAPEL HILL, NORTH CAROLINA
Single-Stage Two-Stage
Operating Operating
Description Conditions Conditions
Flow, gpm 1.2 2.4
Influent BOD, mg/1 179 179
Influent SS, mg/1 247 247
Primary Clarifier
Overflow Rate, gpd/sq ft 470 628
BOD Removal, percent 35 30
Trickling Filter - 1st Stage
Depth, ft 4.25 4.25
Hydraulic Loading, mgd/acrel 18.0 23.9
Organic Loading, Ib BOD/day/1,000 cu ft1 34.4 73.5
Recirculation Ratio 2.0 1.0
Final or Intermediate Clarifier
Overflow Rate, gpd/sq ftl 436 314
BOD Removal - 1st Stage Filter and Clarifier, percent 69 59
Trickling Filter - 2nd Stage
Depth, ft - 4.25
Hydraulic Loading, mgd/acrel 23.9
Organic Loading, Ib BOD/day/1,000 cu ft1 - 29.8
Recirculation Ratio 1.0
Final Clarifier
Overflow Rate, gpd/sq ft - 436
BOD Removal - 2nd Stage Filter and Clarifier, percent ~ 55
SS Removal - 2nd Stage Filter and Clarifier, percent - 44
Overall Plant Performance
BOD Removal, percent 80 87
SS Removal, percent 85 93
Effluent BOD, mg/1 36 23
Effluent SS, mg/1 36 18
1 Includes recirculation.
4-23
-------
FIGURE 4-5
UPGRADING A SINGLE STAGE TRICKLING FILTER
TO A TWO-STAGE BIOLOGICAL FILTRATION/ACTIVATED SLUDGE SYSTEM
INFLUENT
8 1 MGD
PRIMARY TRICKLING FINAL
CLARIFIERS FILTERS CLARIFIERS
PUMP
1
SLUDGE
SETTLED FINAL SLUDGE
EFFLUENT
TREATMENT SYSTEM BEFORE UPGRADING
INFLUENT
6,9 MGD
INTERMEDIATE
CLARIF1ERS-
PRIMARY TRICKLING
CLARIFIERS FILTERS
PUMP
AERATION FINAL
TANKS CLARIFIERS
SLUDGE
/
1
fc-
A
SLUDGE
k
A
RETURN SLUDGE ~p
EFFLUENT
.4 MGD
SLUDGE
PUMP
I
I WASTE
+ SLUDGE
TREATMENT SYSTEM AFTER UPGRADING
4-24
-------
TABLE 4-6
OPERATING DATA FOR KANKAKEE, ILLINOIS
Year Before Year After
Description Upgrading Upgrading
Flow, mgd 8.1 6.9
Influent BOD, mg/1 292 313
Influent SS, mg/1 280 332
Primary Clarifier
Overflow Rate, gpd/sq ft 700 600
BOD Removal, percent 21 47
SS Removal, percent 30 54
Trickling Filter
Depth, ft 6.4 6.4
Hydraulic Loading, mgd/acre 13.3 11.3
Organic Loading, Ib BOD/day/1,000 cu ft 93.4 56.9
BOD Removal, percent 28 27
Final (or Intermediate) Clarifier
Overflow Rate, gpd/ft 1,050 900
BOD Removal - 1st Stage Filter and Clarifier, percent 43 80
SS Removal - 1st Stage Filter and Clarifier, percent 53 79
Complete-Mixed Aeration Tank
Detention Time Based on Average Flow, hr^ 8.2
Sludge Recycle Rate, percent of Average Flow 20
Volumetric Loading, Ib BOD/day/1,000 cu ft
Aeration Tank Volume 25
Final Clarifier
Overflow Rate, gpd/sq ft - 610
BOD Removal - 2nd Stage Filter and Clarifier, percent 77
SS Removal - 2nd Stage Filter and Clarifier, percent 76
Overall Plant Performance
BOD Removal, percent 67 98
SS Removal, percent 73 99
Effluent BOD, mg/1 95 6
Effluent SS, mg/1 75 4
^Includes sludge recycle.
4-25
-------
4.5.1.4 Upgrading an Existing Two-Stage Trickling Filter to a Multiple-Stage
Biological System
Several options are available for upgrading a hydraulically or organically overloaded
two-stage filter. Three of the more common techniques are listed below:
1. Construction of a roughing filter preceding the existing system
2. Construction of an activated sludge system following the existing system
3. Construction of a separate parallel biological treatment system.
A detailed discussion will not be presented here, since most of the engineering
considerations pertaining to these three options have been examined in previous sections.
4.5.2 Upgrading to Increase Organic Removal Efficiency
Upgrading techniques previously discussed relate to the ability of existing facilities to handle
increased hydraulic or organic loads by providing modifications to meet existing effluent
standards. However, there may be a need to meet higher effluent standards even though the
existing facilities are not hydraulically or organically overloaded. Table 4-7 contains
suggested alternatives for improving effluent quality under these conditions. The main
purpose of the table is to present various alternatives and to suggest a range of anticipated
improvement in performance for each alternative.
It should be emphasized that in cases where unit processes are added to existing facilities,
the improvement in overall organic removal will be a direct function of the BOD removal
achieved in the "add-on" unit process. However, where unit processes precede existing units,
e.g., the use of a roughing filter, the overall BOD removal may not be increased in direct
proportion to the amount achieved in the "add-on" process.
A detailed discussion on polishing lagoons, microscreens, filters, activated carbon and
clarifier modifications appears in subsequent chapters. The applicability of alternatives to
individual cases should be evaluated in detail prior to the implementation of a particular
upgrading procedure.
4.5.2.1 Upgrading a Single-Stage Trickling Filter Through Conversion to a
Complete-Mix Activated Sludge System
In 1965, the Ontario Water Resources Commission set 15 mg/1 of BOD and SS as the
objectives for secondary treatment plant effluents. This effluent quality could not be
achieved with an existing high-rate trickling filter plant at Gravenhurst, Ontario (32).
4-26
-------
f"
to
TABLE 4-7
UPGRADING TECHNIQUES FOR IMPROVEMENT OF TRICKLING FILTER PLANT EFFICIENCY
Addition Preceding
Existing Unit
Roughing Trickling Filter
(Rock or Synthetic Media)
Chemical Addition
To Primary Clarifier
Modification to
Existing Unit
1. Low-Rate Trickling Filter
Add recirculation during
low-flow periods
2. High-Rate Trickling Filter
Increase recirculation.
3. Two-Stage Trickling Filterl
Addition Following
Existing Unit
Incremental BOD Removal
Across the Added or
Modified Process
percent
2nd Stage Activated Sludge^
Polishing Lagoon
Multimedia Filters
Microscreening
Activated Carbon
^Generally not amenable to modifications for increasing treatment efficiency.
consideration if year-round nitrification is required.
0-10
0-10
2040
30-50
30-70
30-60
50-80
30-80
60-80
-------
To upgrade the high-rate filter, the plant was converted to complete-mix activated sludge.
The 40-foot diameter filter was converted to an aeration tank by removing the media and
raising the concrete sidewalls 7 feet to a total height of 12 feet. A 10-hp mechanical aerator
was installed. The duo-clarifier (combination primary and secondary clarifier) was converted
to a 40-foot diameter secondary clarifier, and a new 35-foot diameter primary clarifier was
constructed. A 100-percent sludge recycle capacity was provided.
The previously described upgrading technique resulted in the following measured
improvements:
Before After
Parameter Upgrading Upgrading
Dry weather design flow, gpd
Influent organic load, Ib BOD/day
Effluent BOD, mg/1
300,000
360
> 20
375,000
540
15-20
The upgrading technique employed allowed the plant to handle 50 percent greater BOD
loads while producing an effluent of higher quality.
The capital costs for this upgrading were estimated at $82,000 and were allocated as
follows:
Tank modification
Secondary clarifier modification
Total
$64,000
18,000
$82,000
These costs do not include upgrading of any other unit treatment processes, e.g., primary
clarification.
4.5.3 Upgrading Existing Trickling Filters to Provide Nitrification
Increasingly stringent State and Federally approved water quality standards are requiring the
partial or complete removal of nitrogenous oxygen-demanding materials from critical stream
basins, lakes and estuaries. For this reason, many existing trickling filter plants will have to
be upgraded to provide nitrification.
The development and maintenance of nitrifying organisms in trickling filter systems is
mainly dependent upon organic loading and wastewater temperature. Generally,
nitrification occurs best at low BOD loadings and high wastewater temperature (20 deg C or
higher). Table 4-8 clearly shows that the degree of nitrification in trickling filters improves
as the volumetric BOD loading is reduced.
4-28
-------
TABLE 4-8
TRICKLING FILTER NITRIFICATION DATA
f'
tb
Plant
Livermore, Calif.
Glenwood City, Wise.
Lakefield, Minn.
Allentown, Pa.
Ft. Benjamin Harrison, Ind,
Fitchburg, Mass.
Salford, England
Midland, Mich. - Summer
- Winter
Organic
Loading
BOD
lb/day/1,000
cuft
110
66
54
18
4.6
3.7
3.2
4.6
5.9
7.7
9.2
11.8
16.3
22.6
4.2-5.6
4.2-5.6
Influent
BOD
mg/1
50
168
296
209
113
100
206
199
192
165
239
191
235
266
15-20
15-20
Hydraulic
Loading
mgd/acre
3.4
14.3
7.1
4.4
1.7
4.0
0.7
1.0
1.3
2.0
1.6
2.6
2.9
3.6
31
31
Recirculation
Depth Ratio
ft
4.25
7.0
7.5
10.0
8.0
10.0
8.0
8.0
8.0
8.0
8.0
8.0
8.0
8.0
21.5
21.5
2.0
2.0
0.3
0.1
0
0
0/1
0/1
0/1
0/1
0/1
0/1
0/1
0/1
0-1.0
1.0
Influent
NHa - N
mg/1
40.7
10.6
19.0
30.0
21.0
12.0
36.6
38.3
40.7
40.5
43.9
32.0
31.3
33.9
12.7
14.0
Effluent
NH3-N
mg/1
32.6
6.7
16.3
12.0
6.3
2.0
0.7/0.4
2.8/0.9
5.7/2.8
11.4/4.9
12.5/2.2
9.7/4.8
16.9/11.8
19.7/13.6
1.5
2.2
NH3-N
Removal
percent
20
37
13
60
70
83
93/99
93/98
86/93
72/88
72/95
70/85
46/62
42/60
88
84
Reference
33
34
34
35
7
36
37
38
-------
In colder climates, existing trickling filter plants may be modified to achieve a high degree
of nitrification on a year-round basis, or only during the warmer months of the year,
depending on effluent criteria. Year-round nitrification facilities must be designed for the
lowest wastewater temperatures experienced in the winter months. In this instance, the
required filter volume is generally much greater than that required for seasonal nitrification
and normally requires at least two-stage treatment. Figure 4-6 illustrates four possible
systems for implementing year-round nitrification.
Table 4-8 indicates that a high degree of nitrification (> 80 percent) was achieved at
Salford, England (37) at BOD loading rates less than 12 and 6 Ib BOD/day/1,000 cu ft with
recirculation ratios of 1.0 and zero, respectively. These results were obtained with 20-foot
diameter trickling filters, 8 feet deep, with a media consisting of blast furnace slag. At
Midland, Michigan, a pilot study was conducted by Dow Chemical for the U. S. EPA (38),
using synthetic media in a 3-foot diameter by 21.5-foot deep column. Table 4-8 shows that
secondary effluent fed to the pilot facility at organic loading rates less than
5.6 Ib BOD/day/1,000 cu ft produced a high degree of nitrification at temperatures as low
as 7 deg C. The data shown in Table 4-8, and similar operating data from other locations,
indicate that trickling filters can achieve a high degree of nitrification at organic loadings less
than 5 Ib BOD/day/1,000 cu ft.
Example C on Figure 4-6 utilizes an activated sludge system which can also be used as the
second-stage nitrification facility. This scheme can be used with or without an intermediate
clarifier. If plant hydraulics permit, the existing first-stage clarifier may be converted for use
as a final clarifier. The design of this type of facility is discussed in Chapter 5 and in
reference (39).
The following examples describe the upgrading of existing trickling filter plants for
nitrification.
4.5.3.1 Upgrading a Single-Stage Trickling Filter to a Two-Stage Filtration
System to Provide Nitrification
In January 1971, the Pennsylvania Department of Health upgraded the water quality criteria
established for the Lehigh River and its major tributaries. As a result, the City of Allentown
was informed that upgrading of its wastewater treatment plant would be required. The
upgraded plant was designed to meet the following effluent standards from May 1 to
October 31: BOD not to exceed 20 mg/1, ammonia nitrogen (NH3-N) not to exceed 3 mg/1
and total SS not to exceed 30 mg/1. It was also necessary to expand plant capacity from
28.5 to 40 mgd.
The existing plant provided secondary treatment using fixed-nozzle rock-media trickling
filters as shown on Figure 4-7. The raw wastewater to the plant contained 190 mg/1 BOD,
16 mg/1 NH3-N, and 215 mg/1 SS, which was reduced through the plant to effluent
concentrations of 40 mg/1 BOD, 12 mg/1 NH3-N, and 40 mg/1 SS.
4-30
-------
FIGURE 4-6
UPGRADING A TRICKLING FILTER SYSTEM
TO PROVIDE TWO-STAGE NITRIFICATION
EXAMPLE A
FINAL
EFFLUENT
EXAMPLE D
PRIMARY
^
EFFLUENT
AERATION
TANK
fc
CLARIFIER
NOTES' I. CROSS HATCHED FACILITIES DESIGNATE EXISTING TRICKLING
FILTERS AND CLARIFIERS,
2 RECIRCULATION SCHEMES NOT SHOWN
4-31
-------
INFLUENT
24 MGD
FIGURE 4-7
UPGRADING A SINGLE-STAGE TRICKLING FILTER
TO A TWO STAGE FILTRATION SYSTEM
TO PROVIDE NITRIFICATION
PRIMARY
TANKS
TRICKLING
FILTERS
FINAL
CLARIFIERS
k.
i
h
V
1
R
/
SL
r
RECIRCULATION
SLUDGE
TREATMENT SYSTEM BEFORE UPGRADING
EFFLUENT
PRIMARY
TANKS-
SYNTHETIC MEDIA
FIRST-STAGE
FILTERS
INTERMEDIATE
PUMPING STATIONS
INFLUENT
40 MGD
INTERMEDIATE
CLARIFIERS
EXISTING TRICKLING
FILTERS-NITRIFICATION
UNIT-^
-EXPANDED
FINAL CLARIFIERS
EFFLUENT
RECIRCULATION 4.0 MGD
TREATMENT SYSTEM AFTER UPGRADING
SLUDGE
4-32
-------
Under the plan (35) recommended to meet the additional removal and flow requirements,
an auxiliary pumping station was installed near the existing main station. The existing
detritus tank and grit washing facilities were replaced by aerated grit chambers. New
synthetic media first-stage filters were installed. The old rectangular primary tanks were
used as intermediate clarifiers following the first-stage filters and new circular primary
clarifiers were installed. The existing rock-media trickling filters were retained as
nitrification units. The final clarifiers and chlorine contact tanks were expanded and
modifications were made to the existing piping. Recirculation of flow to the trickling filters
will only be used during low flow conditions to maintain a minimum hydraulic wetting rate.
A flow diagram of the upgraded plant is shown on Figure 4-7. Operating conditions and
upgraded design conditions are given in Table 4-9.
The capital costs associated with this upgrading were estimated at $11,632,000 and were
allocated as follows:
Pumping $ 1,192,000
Grit Handling Facilities 423,000
Primary Clarifiers 1,200,000
Trickling Filters 3,985,000
Final Clarifiers 414,000
Chlorine Contact Tank Additions 80,000
Sludge Processing 745,000
Instrumentation, Electrical, Plumbing and HVAC 1,647,000
Sitework, Outside Pumping and Dewatering 1,800,000
Miscellaneous 146,000
Total $11,632,000
4.5.3.2 Upgrading a Single-Stage Trickling Filter to a Two-Stage System to
Provide Nitrification
In 1969, an industry in New York State was required by the Department of Health to
upgrade its existing trickling filter plant to meet year-round effluent standards of about
160 Ib/day of Ultimate Oxygen Demand (ultimate BOD plus nitrogenous oxygen demand).
The plant consisted of single-stage, high-rate trickling filters designed to treat 0.9 mgd of
sanitary and industrial wastes generated by the industry. The treatment system before
upgrading is shown on Figure 4-8.
To meet the more stringent standard, it was decided to follow the existing facility with an
activated sludge nitrification system (40). Except for some modifications in trickling filter
recirculation, the existing facility was incorporated into the upgrading scheme unchanged.
Effluent from the trickling filter plant was fed to a four-compartment, plug-flow
nitrification reactor equipped with mechanical aerators. Two new final clarifiers followed
the nitrification reactor. A flow diagram of the upgraded plant is shown on Figure 4-8.
4-33
-------
TABLE 4-9
OPERATING AND DESIGN CONDITIONS
FOR ALLENTOWN, PENNSYLVANIA
Description
Flow, mgd
Influent BOD, mg/1
Influent SS, mg/1
Influent NH3 - N, mgA
Primary Clarifier
Overflow Rate, gpd/sq ft
BOD Removal, percent
First-Stage Filter - Synthetic Media
Depth, ft
Hydraulic Loading, mgd/acre
Organic Loading, Ib BOD/day/1,000 cu ft
Intermediate Clarifier
Overflow Rate, gpd/sq ft
1st Stage BOD Removal, percent
1st Stage NHg - N Removal, percent
Trickling Filter - Rock Medial
Depth, ft
Hydraulic Loading, mgd/acre
Organic Loading, Ib BOD/day/1,000 cu ft
Final Clarifier
Overflow Rate, gpd/sq ft
2nd Stage BOD Removal, percent
2nd Stage NHg - N Removal, percent
Overall Plant Performance
BOD Removal, percent
SS Removal, percent
NHg - N Removal, percent
Effluent BOD, mg/1
Effluent SS, mg/1
Effluent NH3 - N, mg/1
^Proposed nitrification unit.
2May 1 - October 31.
1971
Operating
Condition
24
190
215
16
870
10
4.5
18.0
760
79
81
25
40
40
12
Upgraded
Design
Condition
40
210
230
15
920
31
32
58
50
1,500
80
0
10
7.5
4.3
800
33
80
91
87
80
202
302
32
4-34
-------
FIGURE 4-8
UPGRADING A SINGLE-STAGE TRICKLING FILTER TO
A TWO-STAGE BIOLOGICAL SYSTEM TO PROVIDE NITRIFICATION
INDUSTRIAL FLOW
SANITARY
FLOW
0. 11 MGD
EFFLUENT
PUMP
STATION
0,55 MGD
PUMP
STATION
TREATMENT SYSTEM BEFORE UPGRADING
INDUSTRIAL FLOW
SANITARY FLOW
0. 25 MGD
PUMP
STATION
EXISTING
TRICKLING^
FILTERS'
INTERMEDIATE
CLARIFIER '
ISTING FINAL
CLARIFIER)
^
, '
NITRI-
FICATION
REACTOR
RETURN SLUDGE
FINAL
CLARIFIERS
i
P
S
EFFLUENT
PUMP
STATION
TREATMENT SYSTEM AFTER UPGRADING
4-35
-------
Operating conditions and upgraded design conditions are given in Table 4-10. The upgraded
facility has been in operation for over a year and has produced an effluent of higher quality
than specified.
The capital cost for the construction of the upgraded facility was estimated at $690,000 in
1970.
4.5.4 Upgrading to Remove Nutrients
The preceding discussion has been primarily concerned with upgrading techniques to
improve treatment plant effluent quality by reducing the oxygen demand that the effluent
will exert on the receiving water. A matter of increasing concern, however, has been the
nutrient content of treatment plant effluent. It is known that the continued and normal
growth of microorganisms requires the availability of certain elements and nutrients as well
as an energy source. Nitrogen and phosphorus are nutrients normally present in substantial
quantities in raw wastewater that are not effectively removed by conventional biological
treatment. The discharge of these substances contributes to the overfertilization and
eutrophication of our surface waters. As a result, many states have established or are
considering standards to limit the discharge of nutrients. The techniques that have been
developed for nutrient removal are summarized in Table 4-11.
The design of a plant for nitrogen removal generally requires separate biological units for
secondary treatment, nitrification and denitrification. In upgrading a trickling filter plant to
remove nitrogen, the existing trickling filters may be used for roughing, as secondary
treatment units, or as nitrification units.
The techniques available for removal of phosphorus are the subject of detailed discussion in
the Process Design Manual for Phosphorus Removal (41). The following case histories
illustrate the upgrading of existing trickling filter plants for phosphorus removal.
4.5.4.1 Upgrading a Low-Rate Trickling Filter System With Chemical Addition
for Phosphorus Removal
A plant-scale study was begun at Richardson, Texas, in 1970 to evaluate the potential of
chemical addition to remove phosphorus from and improve overall performance of the
City's trickling filter plant (5). Major objectives were to reduce the effluent phosphorus
concentration to 1.0 mg/1 (as P), or less, and to reduce effluent BOD and SS residuals to
15 mg/1, or less.
The existing plant was a low-rate trickling filter facility with combination primary
clarifier/digester units (clarigesters) as shown on Figure 4-9. Operating data for a control
period before the addition of chemicals are shown in Table 4-12.
4-36
-------
TABLE 4-10
OPERATIONAL AND DESIGN DATA FOR AN INDUSTRY
IN NEW YORK STATE
Description
Wastewater Flow, mgd
Sanitary
Industrial
Total
Influent BOD, mgA
Sanitary
Industrial
Influent SS, mgA
Sanitary
Industrial
Primary Clarifier
Overflow Rate, gpd/sq ft
BOD Removal (Sanitary Flow), percent
Trickling Filters - Stone Media
Depth, ft
Hydraulic Loading, mgd/acrel
Organic Loading, Ib BOD/day/1,000 cu ft2
Recirculation Ratio
Intermediate Clarifier (Existing Final Clarifier)
Overflow Rate, gpd/sq ftl
1st Stage BOD Removal, percent
Nitrification Reactor
Volume, cu ft
Depth, ft
Volumetric Loading, Ib BOD/day/1,000 cu ft
MLSS, mg/1
F/M, Ib BOD/day/lb MLSS
Sludge Recycle Rate, percent of Average Flow
Oxidizable Nitrogen to Reactor, mgA
1970
Operating
Condition
0.11
0.51
0.62
485
55
224
43
336
30
5
22.6
34
1.67
740
66
Upgraded
Design
Condition
0.25
0.64
0.89
500
120
250
90
435
30
5
20.4
86
0.67
560
50
55,500
12
12.5
4,000
0.05
100
32.5
4-37
-------
TABLE 4-10 (Continued)
Description
1970
Operating
Condition
Upgraded
Design
Condition
Final Clarifier
Overflow Rate, gpd/sq ft
2nd Stage BOD Removal, percent
Overall Plant Performance
BOD Removal, percent
SS Removal, percent
.Effluent BOD, mgA
Effluent SS, mgA
Effluent NH3 - N, mgA
^Includes recirculation.
not include recirculation.
72
44
36
42
23
280
92
97
>90
7
<15
1.4
Nitrogen
TABLE 4-11
SUMMARY OF TREATMENT PROCESSES FOR
NUTRIENT REMOVAL
Physical-Chemical
Ammonia Stripping after
pH Adjustment
Ammonia Removal by Breakpoint
Chlorination
Biological
Nitrification-Denitrification
Algae Harvesting
Phosphorus
Ammonia Removal by Ion Exchange
Chemical Precipitation
4-38
-------
FIGURE 4-9
UPGRADING A STANDARD RATE TRICKLING FILTER SYSTEM
WITH CHEMICAL ADDITION TO PROVIDE PHOSPHORUS REMOVAL
INFLUENT
1,5 MGD
PRIMARY
CLARIFIERS TRICKLING
(CLARIGESTERS) FILTERS
FINAL
CLARIFIER
DIGESTED
I SLUDGE
V TO SAND BEDS
SETTLED FINAL SLUDGE
EFFLUENT
TRICKLING FILTER SYSTEM BEFORE UPGRADING
PRIMARY
CLARIF
(CLAR
INFLUENT
. u MGD t
IERS TRICKLING
GESTERS) FILTERS
f\
^\ ^^"
\ D GESTED
4 SLUDGE
TO SAND BEDS
SETTLED FINAL
ALUM |
V b
INAL
-LARIFIER
SLUDGE
^^^^ ^.
TRICKLING FILTER SYSTEM AFTER UPGRADING
EFFLUENT
4-39
-------
TABLE 4-12
OPERATING DATA FOR RICHARDSON, TEXAS
Description
Flow, mgd
Influent BOD, mg/1
Influent SS, mg/1
Influent P, mg/1
Primary Clarifier
Overflow Rate, gpd/sq ft
Primary Effluent BOD, mg/1
Primary Effluent SS, mg/1
Primary Effluent P, mg/1
BOD Removal, percent
SS Removal, percent
P Removal, percent
Trickling Filter
Depth, ft
Recirculation Ratio
Hydraulic Loading, mgd/acre
Organic Loading, Ib BOD/day/1,000 cu ft
Secondary Clarifier
Alum Dosage Rate, Average A1:P mole ratio
Overflow Rate, gpd/sq ft
Secondary BOD Removal, percent
Secondary SS Removal, percent
Secondary P Removal, percent
Overall Plant Performance
Effluent BOD, mg/1
Effluent SS, mg/1
Effluent P, mg/1
BOD Removal, percent
SS Removal, percent
P Removal, percent
*Not reported.
**Based on Influent P.
1970
Control
Period
1.5
166
155
11
400
*
*
6.5
0
3.9
390
20
15
8
88
90
27
1971-72
Extended
Alum Run
(11-1/2 Months)
1.6
170
155
11.4
425
115
110
8.6
32
29
25
6.5
0
4.1
14.0
1.6**
415
96
94
94
5
7
0.5
97
95
96
4-40
-------
Liquid alum and liquid ferric chloride were selected as the two coagulants to be used in
operational chemical addition trials. Points of chemical injection evaluated were just ahead
of the influent wet well and in the junction box ahead of the final clarifier. The preliminary
trials provided the following observations:
1. Ferric chloride addition at either injection point resulted in about 80 percent
overall phosphorus removal, actually downgraded plant BOD and SS removals and
produced excessive carry-over of discrete iron colloids in the plant effluent.
2. Alum addition to the influent wet well, although somewhat more effective than
ferric chloride addition in achieving the desired effluent quality, resulted in a
rapid and dramatic decrease in digester alkalinity and pH, necessitating
termination of the trial in nine days.
3. Alum addition just ahead of the final clarifier proved to be the most effective
chemical injection technique for removing phosphorus and upgrading BOD and SS
removals, and did not exhibit the deleterious side effect on the unheated digesters
observed when alum was dosed to the raw wastewater.
Based on the above operational trials, alum addition ahead of the final clarifier was selected
as the most promising chemical additive system for an extended 11-1/2-month run. The
flow diagram for the upgraded system is shown on Figure 4-9.
It was found that a mole ratio (Al: Influent P) of 1.5 to 1.7 consistently yielded effluent
concentrations of 5 mg/1 of BOD, 7 mg/1 of SS and 0.5 mg/1 of total phosphorus (as P). The
corresponding effluent values for this plant prior to chemical addition were 20, 15 and
8 mg/1, respectively. The improved plant performance obtained with this upgrading
technique was attributed in part to the low final clarifier overflow rate, careful management
of final clarifier sludge withdrawal to prevent disruption to and loss of the alum floe blanket
and frequent manual adjustment of the chemical feed pump rate to match alum dosage to
mass inflow of phosphorus. Alum treatment doubled the volume of anaerobically digested
sludge produced. However, the digested alum/biological sludge exhibited superior drying
characteristics on sand beds and could be removed in about one-half the normal time.
Operating data for the upgraded plant are shown in Table 4-12.
Chemical costs were $0.05/1,000 gal of plant flow or $0.36/lb of phosphorus removed, with
phosphorus removal at the 96 percent level. The 1970 capital costs associated with plant
modifications for chemical addition were $65,000, allocated as follows:
New laboratory building $21,000
Laboratory equipment and furniture 6,000
Chemical storage and feed equipment, plant piping
and metering modifications 38,000
Total $65,000
4-41
-------
4.5.4.2 Upgrading a High-Rate Trickling Filter System With Chemical Addition
for Phosphorus Removal (A)
Promising results were obtained in removing phosphorus and in generally improving plant
performance at Richardson, Texas, by the addition of alum to trickling filter effluent as
discussed in Subsection 4.5.4.1. In view of these results, the University of North Carolina
Wastewater Research Center initiated a follow-up study to further explore and confirm this
process at the Chapel Hill, North Carolina Wastewater Treatment Plant (7). Conducting a
similar chemical addition project at Chapel Hill was considered worthwhile because it is a
typical high-rate trickling filter plant using recirculation, whereas the Richardson filters were
low-rate units. Furthermore, parallel and identical lines of treatment units were available at
Chapel Hill, allowing direct comparison of results with and without alum addition.
Comparison of parallel results was not possible at Richardson.
The wastewater treatment plant at Chapel Hill is a conventional high-rate installation
treating predominately domestic wastewater. Incoming wastewater passes through
pretreatment facilities for the removal of large solids and grit. The flow is then divided into
equal portions for diversion to two identical lines of treatment, each consisting of a primary
clarifier, trickling filter and final clarifier as shown on Figure 4-10. On one side of the plant,
a feed pump system was installed to add liquid alum to the trickling filter effluent just prior
to the final clarifier.
Operating data for the two sides of the plant are shown in Table 4-13. Significant
improvements in BOD, SS and phosphorus removals were achieved with an alum dosage rate
that ranged between an Al:Influent P mole ratio of 1.5 and 2.2. Within this range and for
the loading rates shown, the final clarifier hydraulic loading appeared to be the significant
factor affecting process efficiency. Subsequent phases of the study indicated that
phosphorus removals of over 90 percent and BOD removals of about 95 percent could be
consistently obtained by lowering the final clarifier overflow rate to 500 gpd/sq ft.
Recirculation of the final clarifier settled alum/humus sludge to the primary clarifier for
thickening had a beneficial effect on primary treatment efficiency and decreased the organic
loading on the alum train trickling filter.
Some problems were encountered at Chapel Hill in the sludge digestion and digester
thickening operations during alum addition (see Section 11.2.6).
4.5.4.3 Upgrading a High-Rate Trickling Filter System With Chemical Addition
for Phosphorus Removal (B)
In 1968, the City of Marlborough, Massachusetts, initiated a series of studies to improve the
performance of its existing high-rate trickling filter plant (42). It was hoped that chemical
treatment of the filter effluent prior to final settling would provide acceptable effluent
concentrations of BOD, SS and phosphorus, with nominal additional capital costs. In 1971,
4-42
-------
following pilot studies, a plant-scale program of alum addition was begun. A flow diagram of
the existing single-stage, high-rate trickling filter system is shown on Figure 4-11.
The existing plant includes Imhoff tanks for primary treatment and sludge digestion.
Secondary humus sludge is returned to the Imhoff tanks' 'influent. Recirculation is
accomplished with either trickling filter or final clarifier effluent. Pre-upgrading operating
data for the period from January, 1970 to April, 1971 are shown in Table 4-14.
The upgrading modifications included the installation of alum storage and feeding
equipment. Operating data for the upgraded plant from May, 1971 through August, 1972
are also shown in Table 4-14. As can be seen, the recycle of alum-laden sludge from the final
clarifier substantially improved the performance of the Imhoff tanks.
The capital cost for the chemical storage and feed system was $4,250. Based on a chemical
cost of $65/ton of dry alum and a dosage of 1.5 moles of Al+3 to 1 mole of inorganic
phosphorus, the treatment cost was $0.32/1,000 gal of plant flow.
4-43
-------
FIGURE 4-10
UPGRADING A TRICKLING FILTER SYSTEM
USING CHEMICAL ADDITION FOR PHOSPHORUS REMOVAL
PRIMARY TRICKLING FINAL
rONTROL CLARIFIER FILTER CLAR
TRAIN - /^~X
lUCMICUT W h/ \ fe
1,4 MGD =
]/2 PLANT
INFLUENT
i "A J
i COMBINED
SLUDGES
i
i W
FIER
fe.
w
CJ
ce.
^
CJ
LU
ce
r
ca
LU LU
_l I C9
1 «r CD
^ ;z "~*
LU _J
CO U_ CO
EFFLUENT
TRICKLING FILTER TRAIN WITHOUT CHEMICAL ADDITION
TEST
TRAIN
INFLUFNT
1 .4 MGD = '
1/0 D 1 1 U T
Z r L AN 1
INFLUENT
PRIMARY
CLARIFIE
b
1 c
1 s
, "tffil",1 T
>c\ i
\J i
OMBINEO «
LUDGES =
C3
ce
ej
LU
ce
FINAI
CLAR
r
FIER
k.
o
UJ LU
_j _i co
1 -tt C3
1 ^ 13
UJ 1
CO U- CO
EFFLUENT
TRICKLING FILTER TRAIN WITH CHEMICAL ADDITION
4-44
-------
TABLE 4-13
OPERATING DATA FOR CHAPEL HILL, NORTH CAROLINA
Side 1 Side 2
Description (No Alum) (With Alum)
Flow, mgd 1.41 1.41
Influent BOD, mg/1 168 168
Influent SS, mg/1 229 229
Influent P, mg/1 11.3 11.3
Primary Clarifier
Overflow Rate, gpd/sq ft 1,100 1,100
Primary Effluent BOD, mg/1 77 63
Primary Effluent SS, mg/1 89 68
Primary Effluent P, mg/1 9.7 6.6
BOD Removal, percent 54 63
SS Removal, percent 61 70
P Removal, percent 14 42
Trickling Filter
Depth, ft 4.25 4.25
Organic Loading, Ib BOD/day/1,000 cu ft 2.0 2.0
Hydraulic Loading, nigd/acre^ 16.3 16.3
Organic Loading, Ib BOD/day/1,000 cu ft 18.8 15.4
Secondary Clarifier
Alum Dosage Rate, Average A1:P mole ratio 1.7^
Overflow Rate, gpd/sq ft 885 885,
Secondary BOD Removal, percent 43 76
Secondary SS Removal, percent 29 53
Secondary P Removal, percent 5 73
Overall Plant Performance
Effluent BOD, mg/1 44 15
Effluent SS, mg/1 63 32
Effluent P, mg/1 9.2 1.8
BOD Removal, percent 74 91
SS Removal, percent 72 86
P Removal, percent 19 84
^Includes recirculation.
^Based on Influent P.
4-45
-------
FIGURE 4-11
UPGRADING A TRICKLING FILTER SYSTEM
TO PROVIDE PHOSPHORUS REMOVAL
IMHOFF
TANKS-y
/
INFLUbNI
9 99 ucn ^ [w
T
SLUDGE
TRICKLING
FILTERS-^
/
r-*T\^
^\J^
RECIRCULATION,
h
r
\n
\
NAL
ARI Fl ERS
fe EFFLUFNT
1
1
. J
TRICKLING FILTER SYSTEM BEFORE UPGRADING
r-IMHOFF TRICKLING
\TANKS FILTERSy
\ SECONDARY SLUDGE/ C
INFLUENT / V
i qn u nn " _h »/ i w
w A
SLUDGE
1
ALUM
RECIRCULATION ,
' ^- -
INAL
LARIFIERS
^
|
1
1
1
_l
TRICKLING FILTER SYSTEM AFTER UPGRADING
4-46
-------
TABLE 4-14
OPERATING DATA FOR MARLBOROUGH, MASSACHUSETTS
Description
Flow, mgd
Influent BOD, mg/1
Influent SS, mg/1
Influent Inorganic P, mg/1
Primary Settling - Imhoff Tanks
Overflow Rate, gpd/sq ft
BOD Removal, percent 1
Trickling Filter
Depth, ft
Hydraulic Loading, mgd/acrel
Organic Loading, Ib BOD/day/1,000 cu ft
Secondary Clarifier
Alum Dosage Rate, Average Al: Inorganic P
mole ratio
Overflow Rate, gpd/sq ft
BOD Removal, percent
Percent SS Removal percent^
Overall Plant Performance^
BOD Removal, percent
SS Removal, percent
Inorganic P Removal, percent
Effluent BOD, mg/1
Effluent SS, mg/1
Effluent Inorganic P, mg/1
^Includes 0.8 mgd recirculation.
^Includes primary removal.
^Includes effluent polishing lagoon.
Operating
Conditions
Prior to
Upgrading
2.22
134
112
7
605
3
6
21.4
74
915
54
71
70
85
14
40
17
6
Operating
Conditions
After
Upgrading
1.99
171
190
8.1
520
22
6
19.1
57
1.5
785
77
78
86
88
81
24
23
1.4
4-47
-------
4.6 References
1. Unpublished data, U. S. EPA Office of Water Programs, Washington, D. C. (September,
1973).
2. McKinney, R., Microbiology for Sanitary Engineers. New York: McGraw Hill Book
Company, Inc. (1962).
3. Sewage Treatment Plant Design. Water Pollution Control Federation Manual of
Practice No. 8, Washington, D. C. (1959).
4. Dreier, D.E. Experience in the Operation of Standard Trickling Filters. Sewage Works
Journal (July, 1946).
5. Derrington, R.E., Stevens, D.H., Laughlin, J.E., Enhancing Trickling Filter Plant
Performance by Chemical Precipitation. Environmental Protection Technology Series,
U. S. EPA-670/2-73-060 (August, 1973).
6. Rankin, R.S., Evaluation of the Performance of Biofiltration Plants. Transactions of
the American Society of Civil Engineers, 120, pp. 823-835 (1955).
7. Brown, J.C., Alum Treatment of High-Rate Trickling Filter Effluent, Chapel Hill,
North Carolina, Technology Transfer Design Seminar. Presented at Newark, New
Jersey (March 13-15, 1974).
8. Sewage Treatment at Military Installations. National Research Council, Sewage Works
Journal, 18, No. 5, pp. 787-1,028 (1946).
X*
upf"
9. Recommended Standards for Sewage Works. Great Lakes-Upper Mississippi River
Board of State Sanitary Engineers (1971).
10. Velz, C.J., A Basic Law for the Performance of Biological Beds. Sewage Works Journal,
20, No. 3, pp. 245-261 (1960).
11. Caller, W.S., and Gotaas, H.B., Analysis of Biological Filter Variables. Journal of the
Sanitary Engineering Division, ASCE, 90, No. 6, pp. 59-79 (1964).
12. Schulze, K.L., Load and Efficiency of Trickling Filters. Journal Water Pollution
Control Federation, 32, No. 3, pp. 245-261 (1960).
13. Eckenfelder, W.W., Trickling Filter Design and Performance. Transactions of the
American Society of Civil Engineers, 128, Part III, pp. 371-398 (1963).
4-48
-------
14. Eckenfelder, W.W., and Barnhart, W., Performance of a High-Rate Trickling Filter
Using Selected Media. Journal Water Pollution Control Federation, 35, No. 12, pp.
1,535-1,551 (1963).
15. Bethlehem, Pa.: Private communication with William Grim, Plant Operator
(November, 1970).
16. Burgess, F.J., et al, Evaluation Criteria for Deep Trickling Filters. Journal Water
Pollution Control Federation, 33, No. 8, pp. 787-816 (1961).
17. Deeds and Data. Journal Water Pollution Control Federation, 31, No. 3, pp. 315-320
(1959).
18. Homack, P., Discussion of Article by R. Rankin. Transactions of the American Society
of Civil Engineers, 120, pp. 836-841 (1955).
19. McCabe, J., and Eckenfelder, W., Biological Treatment of Sewage and Industrial
Wastes. New York: Reinhold Publishing Company (1956).
20. Hanumanulu, V., Effect of Recirculation of Deep Trickling Filter Performance. Journal
Water Pollution Control Federation, 41, No. 10, pp. 1,803-1,806 (1969).
21. Ordon, C., Discussion of Article by Baker and Graves (February 1968). Journal of the
Sanitary Engineering Division, ASCE, 94, No. 3, pp. 579-583 (1968).
22. Maier, W., et al, Simulation of the Trickling Filter Process. Journal of the Sanitary
Engineering Division, ASCE, 93, No. 4, pp. 91-112 (1967).
23. Reynolds, L.B., and Chipperfield, P.N.J., Principles Governing the Selection of Plastic
Media for High-Rate Biological Filtration. Presented at the International Congress on
Industrial Waste Water, Stockholm, Sweden (1970).
24. Gulp, G., Direct Recirculation of High-Rate Trickling Filter Effluent. Journal Water
Pollution Control Federation, 35, No. 6, pp. 742-747 (1963).
25. Rowland, W.E., Flow Over Porous Media as in a Trickling Filter. Proceedings12th
Purdue Industrial Waste Conference, Lafayette, Indiana, pp. 435-465 (1957).
26. Eckenfelder, W.W., Industrial Water Pollution Control. New York: McGraw-Hill Book
Company (1966).
27. Benzie, W., Effects of Climatic and Loading Factors on Trickling Filter Performance.
Journal Water Pollution Control Federation, 35, No. 4, pp. 445-455 (1963).
4-49
-------
28. Sheahan, J.P., Use of Styrofoam for Trickling Filter Covers. Proceedings20th Purdue
Industrial Waste Conference, Lafayette, Indiana, pp. 572-582 (1965).
29. Sellards & Grigg, Inc., Sanitary Sewerage and Wastewater Treatment Facilities. An
Engineering Report for the City of Pueblo, Colorado (November, 1971).
30. Brown, J.C., Little, W., Francisco, D.E., and Lamb, J.C., Methods for Improvement of
Trickling Filter Plant Performance, Part I, Mechanical and Biological Optima,
Environmental Protection Technology Series, U. S. EPA Contract No. 14-12-505.
31. Pritchett, J., Kankakee Sewage Treatment Plant. A report on the original facilities and
expansion programs at Kankakee, Illinois (June, 1971).
32. Economical Sewage Treatment Plant Conversion at Gravenhurst. Water and Pollution
Control, 106, No. 1, pp. 26-27 (1968).
33. Hazen and Sawyer Engineers, Upgrading Existing Wastewater Treatment Facilities.
Prepared for U. S. EPA Technology Transfer Design Seminar, Pittsburgh, Pennsylvania
(August, 1972).
34. Walton, Grantham, High-Rate Trickling Filter Performance. Under the direction of the
Board of State Health Commissioners Upper Mississippi River Basin Sanitation
Agreement (March, 1943).
35. Metcalf & Eddy, Inc., Report on The Design for Increased Capacity and Tertiary
Treatment at the Allentown Wastewater Treatment Plant. Allentown, Pennsylvania
(November, 1971).
36. Town Report. Fitchburg, Massachusetts (1928).
37. Stones, T., Investigations on Biological Filtration at Salford. Journal of the Institute of
Sewage Purification, No. 5, pp. 406-417 (1961).
38. Duddles, G.A., and Stevens, E.R., Application of Plastic Media Trickling Filters for
Biological Nitrification Systems. Environmental Protection Technology Series, U. S.
EPA Contract No. 14-12-900 (June, 1973).
39. Nitrification and Denitrification Facilities, Technology Transfer Seminar Publication,
U. S. EPA, Washington, D. C. (August, 1973).
40. Metcalf & Eddy, Inc., Report to (Unnamed Industry) on Sewage and Industrial Waste
Treatment (August 26, 1970).
4-50
-------
41. Process Design Manual for Phosphorus Removal. U. S. EPA, Office of Technology
Transfer, Washington, D.C. (1974).
42. Metcalf & Eddy, Inc., Modifications to the High-Rate Trickling Filter Process. A report
to the Commonwealth of Massachusetts Water Resources Commission (October, 1970).
4-51
-------
-------
CHAPTER 5
TECHNIQUES FOR UPGRADING ACTIVATED SLUDGE PLANTS
5.1 General
A better understanding of the activated sludge process has evolved down through the years,
primarily due to an improved knowledge of the theory involved and to the experience
accumulated in the successful operation of the process. For these reasons, it has become the
most versatile biological treatment process available to the design engineer.
Historically, the activated sludge process has been used in larger cities, where the ratio of
river assimilative capacity to waste load is small. More recently, there has been a trend
toward its use by smaller communities to meet the more stringent requirements of
regulatory agencies.
Existing overloaded conventional activated sludge plants pose a problem to the maintenance
of established water quality standards. Various modifications of the conventional process
developed over the years permit reduced aeration detention time and higher volumetric
loadings. The applicability of these process modifications in the efficient upgrading of
existing plants will be examined and discussed in this chapter.
5.2 Activated Sludge Processes
Basically, the activated sludge process uses microorganisms in suspension to oxidize soluble
and colloidal organics to C02 and H20 in the presence of molecular oxygen. During the
oxidation process, a portion of the organic material is synthesized into new cells. A part of
the synthesized cells then undergoes auto-oxidation in the aeration tank, the remainder
forming excess sludge. Oxygen is required in the process to support the oxidation and
synthesis reactions. To operate the process on a continuous basis, the solids generated must
be separated in a clarifier with the major fraction being recycled to the aeration tank and
the excess sludge being withdrawn from the clarifier underflow for further handling and
disposal.
Due to the adaptability of the process, an activated sludge plant may be designed as a:
1. Conventional Plant
2. Step Aeration Plant
3. Contact Stabilization Plant
4. Complete Mix Plant
5. Modified Aeration Plant
6. Two-Stage Plant
7. Pure Oxygen Activated Sludge Plant.
5-1
-------
It is possible to design a plant which can be operated under several of the above
modifications.
When compared with one another, these plants each have advantages and disadvantages.
Some achieve better BOD and SS removals than others, some cost less to construct, others
cost less to operate, some produce less sludge and some obtain better nutrient removal. All
of these factors must be considered in selecting a particular modification for use in a specific
upgrading situation.
5.2.1 Conventional Activated Sludge
Design of a conventional activated sludge system is usually based upon volumetric loadings
of 20 to 40 Ib BOD/day/1,000 cu ft and organic loadings (food to microorganisms, F/M) of
0.2 to 0.4 Ib BOD/day/lb MLSS. Sludge retention times (SRT) are normally between 5 and
15 days for this process which will achieve 85 to 95 percent BOD removals with proper
operation (1). SRT as used herein is defined as:
Ib MLSS under aeration
(Ib SS wasted + Ib SS lost in final effluent)/day
A schematic for a conventional activated sludge plant is shown on Figure 5-1. The
wastewater is commonly aerated for a period of 6 to 8 hours (based on the average design
flow) in the presence of a portion of the secondary sludge (2). The rate of sludge return
expressed as a percentage of the average wastewater design flow is normally about
25 percent, with minimum and maximum rates of 15 to 75 percent. A plug flow
configuration is achieved in rectangular tanks, designed so that the total tank length is
generally 5 to 50 times the width. Operating data from conventional activated sludge plants
are summarized in Table 5-1.
The following factors have been recognized as limitations in the design and use of a
conventional activated sludge plant:
1. Volumetric BOD loadings are limited to about 40 lb/day/1,000 cu ft because of
poor load distribution.
2. Required aeration detention times are in the range of 6 to 8 hours.
3. A high initial oxygen demand is experienced in the head end of the aeration tank;
however, this can be offset by tapering the air supply.
4. There may be a lack of operational stability with extreme variations in hydraulic
and organic loadings.
5-2
-------
TABLE 5-1
OPERATING DATA FROM CONVENTIONAL ACTIVATED SLUDGE PLANTS
BOD
Plant
Location
Michigan
Illinois
Ohio
Indiana
Oi
CO Maryland
Michigan
Wisconsin
Indiana
Maryland
California
Pennsylvania
Illinois
Influent Sludj
Flow Recy
mgd perce
5.0 32
288.0 48
86.9 25
14.9 30
3.9 32
8.0 16
7.6 52
3.9 31
5.5 29
8.0 26
7.7 25
47.0 45
48.0 34
3.44 20
2.71 26
200.1 38
1 Excluding sludge recycle.
Secondary
Influent
mg/1
182
129
92
161
254
118
157
134
113
155
148
157
181
175
161
119
Secondary
Effluent
mg/1
19
11
12
14
33
6
36
14
6
10
15
6
8
20
14
13
Aeration
Tank
MLSS
mg/1
1,844
1,930
2,180
2,420
1,808
2,801
1,094
2,625
1,680
2,040
2,240
2,449
2,111
1,180
1,160
2,775
Organic
Loading
Ib BOD/day
Ib MLSS
0.34
0.18
0.13
0.16
0.39
0.15
0.39
0.22
0.20
0.23
0.20
0.19
0.23
0.60
0.45
0.17
Volumetric
Loading
Ib BOD/day
l,000cuft
39
21
17
24
44
26
26
35
21
29
25
28
29
45
32
30
Aeration
Detention
Time1
hours
7.0
8.7
7.7
10.0
8.8
6.7
9.1
5.7
8.2
7.7
8.2
8.4
9.2
5.9
7.5
5.8
Air Supplied
per Ib of
BOD Removed
cuft
770
876
1,600
733
500
690
690
886
435
1,260
1,900
1,581
1,352
1,430
1,650
676
Secondary
BOD Removal
Efficiency
percent
90
92
87
91
87
95
77
90
95
94
90
96
96
89
91
89
Reference
3
3
3
3
3
3
3
3
4
5
6
7
-------
FIGURE 5-1
CONVENTIONAL ACTIVATED SLUDGE PLANT
RAW
I/ASTEWATER
PRIMARY
CURIFIER
i
SLUDGE
^
AERATION ~N
^-T»NK +S
^ ^
c ^
w
FINAL
CLARIFIED
RETURN SLUDGE
EXCESS SLUDGE
EFFLUENT
Some of these limitations have stimulated the development and use of various process
modifications, such as step aeration, contact stabilization, complete mix, modified aeration,
two-stage activated sludge and the use of oxygen instead of air as a source of DO. These
process modifications are discussed in subsequent sections.
5.2.2 Step Aeration
A typical flow diagram of a step aeration plant is illustrated on Figure 5-2. Unlike the
conventional plant, the influent wastewater is introduced at several points along the aeration
tank. However, the return sludge, which normally ranges from 25 to 75 percent of the
average design flow, is introduced at the head end of the aeration tank as in the
conventional system. Distributing the influent flow along the aeration tank reduces the
initial oxygen demand usually experienced in the conventional plant. This permits a more
efficient utilization of the activated sludge biomass.
Step aeration plants are usually designed for volumetric loadings of 40 to 60 Ib
BOD/day/1,000 cu ft at F/M's varying from 0.2 to 0.4 Ib BOD/day/lb MLSS. SRT's are
similar to those of the conventional system. A step aeration system will achieve 85 to
95 percent BOD removal. Operating data from step aeration plants are summarized in
Table 5-2.
5-4
-------
TABLE 5-2
OPERATING DATA FROM STEP AERATION ACTIVATED SLUDGE PLANTS
ROD
Plant
Location
New York
New York
New York
New York
New York
New York
Maryland
Indiana
Indiana
Indiana
Pennsylvania
Connecticut
Ontario, Canada
Influent
Flow
mgd
110.0
20.7
92.0
50.0
95.0
31.0
16.9
12.8
19.3
34.3
178.3
37.7
183.0
Sludge
Recycle
percent
24
49
35
28 '
28
28
24
92
50
52
28
34
16
Secondary
Influent
mg/1
74
137
100
120
115
100
140
124
139
131
87
121
115
Secondary
Effluent
mg/1
12
3
8
6
16
12
11
15
17
18
12
17
11
Aeration
Tank
MLSS
mg/1
1,170
3,520
1,110
3,300
3,300
4,400
2,120
2,900
2,750
3,360
2,780
2,540
1,500
Organic
Loading
Ib BOD/day
Ib MLSS
0.49
0.10
0.42
0.31
0.28
0.13
0.54
0.19
0.22
0.22
0.23
0.27
0.40
Volumetric
Loading
Ib BOD/day
l,000cuft
36
23
30
71
58
37
58
33
41
45
40
43
38
Air
Supplied
cu ft/gal
_
-
_
0.43
0.54
0.59
_
_
_
0.58
2.05
1.6
Air Supplied
per Ib of
BOD Removed
cuft
910
910
933
_
_
-
_
1,240
1,080
911
927
2,353
1,580
Aeration
Detention
Time'
hours
3.1
8.4
4.9
2.5
2.9
4.2
3.8
5.3
5.0
4.3
3.2
4.3
4.5
Secondary
BOD Removal
percent
84
94
92
94
86
90
92
89
88
86
86
86
90
Reference
8
8
8
8
8
8
4
8
8
8
9
10
11
1 Excluding sludge recycle.
-------
FIGURE 5-2
STEP AERATION PLANT
AERATION TANK
RAW
WASTEWATER
EFFLUENT
SLUDGE
EXCESS SLUDGE
Biological oxygen requirements for step aeration are similar to those of conventional
activated sludge. However, because of more uniform load distribution, the actual air
supplied is more effectively utilized and the quantity may be somewhat reduced. Since the
detention times are lower than for the conventional system, the air piping and diffusion
equipment must be modified to supply approximately the conventional volume of air to a
tank approximately one-half the conventional size.
In the conventional plant, the MLSS concentration is relatively constant throughout the
aeration tank; in the step aeration plant, the MLSS concentration decreases at each point of
influent addition. The reason that a step aeration system can be operated at the same F/M as
a conventional system, but in about one-half to two-thirds of the aeration tankage, is the
higher average MLSS concentration gained by the incremental addition of flow. The step
feed concept permits the utilization of the same size clarifiers as the conventional system
because the aerator effluent SS concentrations are similar. This principle is shown on Figure
5-3. The concept can also be used to decrease the final clarifier solids loading while
maintaining the same average aerator MLSS concentration.
5.2.3 Contact Stabilization
The principles of the contact stabilization modification were initially demonstrated in the
upgrading of an existing hydraulically overloaded conventional plant (12). As in the step
5-6
-------
FIGURE 5-3
COMPARISON OF SOLIDS LOADING ON THE FINAL CLARIFIER
FOR CONVENTIONAL AND STEP AERATION PLANTS
MODE 1 - CONVENTIONAL
1
25% RETURN
SLUDGE
SS = 10,000 mg/|
100% PRIMARY EFFLUENT
r
A
2,000
B
2,000
C
2,000
D
2,000
AERATOR
EFFLUENT
SS = 2,000 mg/l
AVERAGE AERATOR MLSS CONCENTRATION
2,000 mg/l
MODE 2 - STEP AERATION 25% PRIMARY EFFLUENT/PASS
1
25% RETURN
SLUDGE
k.
SS = 10,000 mg/l
r i
A
5,000
> i
B
3,333
r 1
C
2,500
r
D
2,000
AERATOR
EFFLUENT
SS = 2,000 mg/l
AVERAGE AERATOR MLSS CONCENTRATION
3,208 mg/l
NOTE: EXAMPLE ASSUMES NEGLIGIBLE SS IN PRIMARY EFFLUENT
5-7
-------
aeration process, this modification involves a change in the feed location to the aeration
tanks. Volumetric BOD loadings, F/M, SRT and removal efficiency are similar to those of
the step aeration system. Sludge return ratios vary from 25 to 100 percent of the average
design flow. A schematic for the contact stabilization process is shown on Figure 5-4.
FIGURE 5-4
CONTACT STABILIZATION PLANT
RAW
WASTEWATER
REAERATION
TANK
CONTACT
TANK
RETURN
EFFLUENT
EXCESS
SLUDGE
SLUDGE
Laboratory studies and field work have demonstrated that wastewater BOD in the colloidal
or insoluble state is rapidly removed from wastewater in a relatively short contact time by
the combined mechanism of biological sorption, synthesis and flocculation. This may offer
the possibility of a reduction in plant volume for wastewaters exhibiting these
characteristics. In the contact stabilization process, after the biological sludge is separated
from the wastewater in the clarifier, the concentrated sludge is separately aerated in a
reaeration tank. Here the flocculated and absorbed BOD is stabilized. In addition to a
smaller total aeration volume than required with the conventional activated sludge process,
the contact stabilization process has the advantage of being able to handle greater shock and
toxic loadings because of the biological buffering capacity of the reaeration tank, and the
fact that at any given time the majority of the activated sludge is isolated from the main
stream of the plant flow. Operating data from contact stabilization plants are summarized in
Table 5-3.
The detention times required in the sludge reaeration and contact tanks are interdependent.
The contact tank detention time also depends on wastewater characteristics. For domestic
wastewaters containing normal amounts of insoluble and colloidal BOD, contact tank
5-8
-------
TABLE 5-3
OPERATING DATA FROM CONTACT STABILIZATION ACTIVATED SLUDGE PLANTS
Plant
Location
Texas4
New Jersey
Influent
Flow
mgd
6.3
8.3
7.6
7.6
8.6
6.7
6.5
2.5
Sludge
Recycle
percent
46
35
39
39
39
55
56
70
BOD
Influent
mg/1
330
267
300
331
299
354
358
3125
Effluent
mg/1
22
17
21
19
18
23
21
30
Contact
Tank
MLSS
mg/1
2,321
1,894
1,698
1,377
1,829
2,432
2,533
4,000
Contact
Tank
Aeration
Detention
Timel
minutes
61
46
51
51
45
58
59
100
Reaeration
Tank
MLSS
mg/1
7,072
8,018
7,266
6,050
6,084
6,930
6,917
6,700
Reaeration
Tank
Aeration
Detention
Time2
minutes
392
391
383
383
338
308
312
144
Organic
Loading ^
Ib BOD/day
Ib MLSS
0.34
0.32
0.36
0.49
0.49
0.39
0.38
0.32
Volumetric
Loading **
Ib BOD/day
l,000cuft
123
131
135
149
152
140
138
104
Secondary
BOD Removal
Efficiency
percent
93
94
93
94
94
93
94
90
Reference
12
13
Based on influent flow excluding sludge recycle.
Based on sludge recycle flow.
Based on contact and reaeration volume.
No primary treatment.
Raw wastewater. Organic and volumetric loading computed assuming 25 percent BOD removal in primary clarifiers.
-------
detention times of about 0.5 to 1 hour are employed based on average plant flow, with
reaeration times of 2 to 4 hours based on sludge recycle flow. Total air requirements for this
process are similar to those of conventional activated sludge and are normally divided
equally between the contact and reaeration tanks.
Most of the benefits of contact stabilization are achieved if the organic load is present
mainly in the colloidal state. Generally, the greater the fraction of soluble BOD, the greater
the required contact time. As a result, the required total aeration volume of this process
approaches that of the conventional process as the relative amount of soluble BOD in the
wastewater increases.
Ten-States Standards require significantly higher contact and reaeration times than those
previously cited, especially for smaller sized plants as indicated in Table 5-4 (14). These
standards also specify an F/M of 0.2 to 0.5 Ib BOD/dayAb MLSS and a volumetric loading
of 30 to 50 Ib BOD/day/1,000 cu ft. Although these values were, no doubt, selected to
compensate for the extreme flow variations that occur at small plants, their use may result
in poor quality effluents (15). McKinney (16) has indicated that in typical conservatively
designed contact stabilization plants, all of the stabilization of the organic matter in the raw
wastewater occurs in the contact zone; therefore, only endogenous respiration occurs in the
reaeration tank. This situation often results in poor sludge settling characteristics in the
secondary clarifier.
TABLE 5-4
SUGGESTED DESIGN GUIDELINES (14)
Plant Design
Flow Contact Time! Reaeration Time^ Aerator Loading^
mgd hours hours Ib BOD/day
l,000cuft
to 0.5 3.0 6.0 30
0.5 to 1.5 3.0 to 2.0 6.0 to 4.0 30 to 50
1.6 and up 2.0 to 1.5 4.0 to 3.0 50
* Based on average design flow.
2 Based on average sludge recycle flow.
3 Based on total aeration capacity (reaeration plus contact).
5-10
-------
5.2.4 Complete Mix
In the complete mix process, influent wastewater and recycled sludge are introduced
uniformly throughout the aeration tank, as indicated on Figure 5-5. This flow distribution
results in a uniform oxygen demand throughout the aeration tank which adds some
operational stability when treating slug loads of industrial wastes (17). This process may be
loaded to levels comparable to those of the step aeration and contact stabilization processes
with only slight reductions from the removal efficiencies of those processes. The reduced
efficiency occurs because there is a small amount of short circuiting in the complete mix
aeration tank. Operating data from plants utilizing complete mix are presented in Table 5-5.
5.2.5 Modified Aeration
The flow scheme for modified aeration is similar to that of conventional activated sludge
(Figure 5-1) with the exception that this process is frequently used without primary settling.
In this system, organisms are brought into contact with the incoming wastes for a brief
period of sorption and synthesis. Modified aeration plants are designed to process
volumetric BOD loadings varying from 75 to 150 Ib BOD/day/1,000 cu ft, and will achieve
60 to 75 percent BOD removals under these loading conditions. F/M's normally range from
1.5 to 5.0 Ib BOD/day/lb MLSS, and SRT's vary from 0.2 to 0.5 day. For proper operation,
the return sludge rate normally ranges from 5 to 15 percent of the influent flow (1) (24).
Aeration requirements are less than those of the conventional system due to the reduced
biological activity.
Modified aeration has been proposed as the first stage in a two-stage aeration system where
the second stage will be used for nitrification. Two advantages of using modified aeration in
the first stage are a substantial cost savings in aeration facilities and an SRT sufficiently low
to preclude nitrification. The absence of nitrification and the partial concentration damping
afforded by mixing in the modified aeration tank results in a more uniform ammonia
concentration in the influent to the nitrification system (1) (25).
When modified aeration is used as the first stage of a two-stage system, the first-stage
effluent BOD should be sufficiently high to permit good sludge flocculation and settling in
the second-stage clarifier, but not so high as to cause a high heterotroph growth rate and a
subsequent washout of the nitrifying population in the second stage. The upper limit will
depend on the BOD/NH4 of the second-stage influent. At the U. S. EPA District of
Columbia Pilot Plant, addition of alum helped to maintain the BOD of the first-stage
effluent within the desired range of 30-60 mg/1 (25). Operation of the full scale District of
Columbia Wastewater Treatment Plant has also shown that oxygen requirements for the
modified process are significantly less than for other activated sludge systems and that the
sludge exhibits good settling qualities (24) (25).
Operating data from modified aeration plants are shown in Table 5-6.
5-11
-------
FIGURE 5-5
COMPLETE MIX PLANT
RAW
WASTEWATER OR
PRIMARY EFFLUENT
T fttftt
AERATION TANK
, i i A i 11 11
RETURN SLUDGE
EFFLUENT
EXCESS SLUDGE
5-12
-------
TABLE 5-5
OPERATING DATA FROM COMPLETE MIX ACTIVATED SLUDGE PLANTS
CO
BOD
Plant
Location
Illinois
Minnesota
Nebraska
Nebraska
Nebraska
Texas
Influent
Flow
mgd
1.6
1.94
1.91
1.55
9.8
4.1
4.3
4.6
4.2
4.8
5.8
5.8
3.4
4.1
5.0
0.38
0.43
0.29
0.29
0.30
0.37
Sludge
Recycle
percent
21
21
25
25
158
87
66
62
65
64
37
49
50
100
200
26
40
82
100
145
100
Secondary
Influent
mg/1
102
80
80
108
177
260
270
290
300
350
240
105
250
270
280
225
227
115
141
123
180
Secondary
Effluent
mg/1
8
13
19
18
15
26
16
17
22
34
37
14
15
13.5
6
25
32
9
25
19
17
Aeration
Tank
MLSS
mg/1
6,500
6,000
6,500
6,300
3,750
4,400
4,460
3,920
4,020
4,280
4,040
3,400
4,500
4,500
4,500
4,230
5,460
3,820
5,000
5,540
5,620
Organic
Loading
Ib BOD/day
Ib MLSS
0.17
0.20
0.18
0.20
0.31
0.27
0.32
0.43
0.41
0.51
0.49
0.24
0.27
0.32
0.38
0.48
0.42
0.21
0.20
0.16
0.29
Volumetric
Loading
Ib BOD/day
1,000 cu ft
74
73
72
79
73
80
97
120
110
132
114
50
80
97
116
126
142
50
62
54
103
Aeration
Detention
Time1
hours
2.2
1.8
1.8
2.2
3.6
4.5
4.3
4.0
4.4
3.8
3.2
3.2
5.0
4.4
3.8
2.6
2.5
3.7
3.7
2.2
3.0
Air Supplied
per Ib of
BOD Removed
cuft/lb
1,670
1,900
1,380
1,290
- 2
540
450
500
480
560
570
900
500
500
560
-
-
-
-
-
-
Secondary
BOD Removal
Efficiency
percent
92
84
76
83
92
90
94
94
93
90
85
87
94
95
98
89
86
92
82
85
91
Reference
18
19
20
21
22
23
Excluding sludge recycle.
Mechanical aerators.
-------
TABLE 5-6
OPERATING DATA FROM MODIFIED AERATION ACTIVATED SLUDGE PLANTS
Ul
BOD
Location
New York
Florida
Washington, D. C.
Influent
Flow
mgd
152.0
39.8
42.0
55.8
49.6
53.3
59.2
269.0
273.0
276.0
277.0
Sludge
Recycle
percent
20
8
7
6
6
8
10
12
12
13
11
Secondary
Influent
mg/1
202
205
145
125
165
175
185
154
163
151
146
Secondary
Effluent
mg/1
24
62
59
38
66
62
62
49
44
51
39
Aeration
Tank
MLSS
mg/1
2,000
360
305
275
345
310
430
606
704
775
714
Organic
Loading
Ib BOD/day
IbMLSS
0.9
5.1
4.6
6.2
4.2
4.9
4.3
2.8
2.7
2.3
3.3
Volumetric
Loading
Ib BOD/day
l.OOOcuft
106
138
93
125
127
107
126
105
121
113
134
Air Supplied
per Ib of
BOD Removed
cuft
-
805
1,171
992
1,065
1,061
858
669
739
863
639
Air
Supplied
cu ft/gal
_
0.96
0.84
0.72
0.88
1.0
0.88
0.53
0.68
0.70
0.52
Aeration
Detention
Timel
hours
2.6
2.3
2.2
1.7
2.5
2.3
2.0
2.1
2.0
2.3
1.5
Secondary
BOD Removal
Efficiency
percent
88
68
59
69
60
64
66
68
73
66
73
Reference
26
27
24
Excluding sludge recycle.
-------
5.2.6 Two-Stage Activated Sludge
A two-stage activated sludge plant is essentially two separate activated sludge processes
operating in series, as shown on Figure 5-6. The two separate sludge systems permit the
development of two specialized microbial populations. In the first stage, the bulk of the
carbonaceous material is removed by a wide variety of heterotrophic organisms commonly
found in activated sludge. The reduction of BOD in the first stage permits an accumulation
of the slower growing nitrifying organisms in the second stage which oxidize the ammonia
nitrogen to the nitrate form.
The operating data presented in Table 5-7 indicate that the second-stage effluent BOD is not
sufficiently better than for alternative single stage systems previously discussed to warrant
consideration of this approach unless nitrification is also required. The advantage of
satisfying the oxygen demand of the ammonia nitrogen normally discharged should not be
underestimated since this can amount to as much as 70 percent of the total oxygen demand
of the plant secondary effluent (30). The high air requirement in Ib air supplied/lb BOD
removed in the second stage includes the air used for ammonia oxidation.
5.2.7 Pure Oxygen Activated Sludge
The use of pure oxygen for activated sludge treatment has become competitive with the use
of air due to the development of efficient oxygen dissolution systems. Efficient utilization
of oxygen can be achieved by two methods:
1. Oxygenation is performed in a staged, covered reactor in which oxygen gas is
recirculated within the system until it reaches a level of impurity at which it can
no longer be used. This method is presently used in several municipal and
industrial treatment plants, and is shown schematically on Figure 5-7 (31).
2. Oxygenation is performed in an open reactor in which extremely fine diffusers are
utilized to develop small oxygen gas bubbles that are completely dissolved before
breaking surface in normal-depth tanks (31). This approach has not yet been
implemented in full scale treatment plants.
The use of pure oxygen has the advantages of reduced reactor volume, high effluent DO and
effective odor control. Possible additional advantages include less waste sludge production
and higher waste sludge concentrations. Disadvantages include the increased complexity of
operation of both oxygen generation and dissolution systems and excessive pH depression in
low alkalinity wastewaters when alum is added for phosphorus removal or when nitrification
occurs.
Oxygen gas can be produced at the plant site by either a cryogenic unit or, in the case of
smaller plants, a molecular sieve device. A liquid oxygen storage tank is generally
5-15
-------
FIGURE 5-6
TWO-STAGE ACTIVATED SLUDGE PLANT
FIRST-STAGE
AERATION TANK
RAW
WASTEWATER
OR PRIMARY
EFFLUENT
SECOND-STAGE
AERATION TANK
EFFLUENT
EXCESS SLUDGE
EXCESS SLUDGE
5-16
-------
TABLE 5-7
OPERATING DATA FROM TWO-STAGE ACTIVATED STJJDGE PLANTS
Ist-STAGE PERFORMANCE
Plant
Location
Pennsylvania
Pennsylvania
Ol
1 '
-J
Sludge
Recycle
percent
24
11
23
29
14
19
19
19
19
19
19
BOD
Influent Sludge Ist-Stage Ist-Stage
Flow Recycle Influent Effluent
mgd percent
2.27 27
2.24 19
1.47 48
1.28 46
2.10 27
0.21 56
0.21 56
0.21 56
0.21 56
0.21 56
0.21 56
BOD
2nd-Stage 2nd-Stage
Influent Effluent
mg/1
36
23
32
33
41
12
29
19
25
13
10
mg/1
12
11
19
19
17
7
8
4
21
8
7
mg/1
204
220
271
249
223
138
266
104
133
110
mg/1
36
23
32
33
41
12
29
19
25
13
Aeration
Tank
MLSS
mg/1
3,760
2,020
2,350
1,980
1,920
3,150
3,150
2,650
2,650
2,650
134 10 2,650
2nd-STAGE PERFORMANCE
Aeration
Tank
MLSS
mg/1
1,520
1,600
820
800
935
1,500
1,510
1,350
1,350
1,350
1,350
Organic
Loading
Ib BOD/day
Ib MLSS
0.34
0.21
0.36
0.34
0.59
0.21
0.47
0.35
0.49
0.24
0.20
Volumetric
Loading^
Ib BOD/day
1,000 cuff
33.0
21.0
19.0
17.0
35.0
20.5
44.0
29.0
41.5
20.5
17.0
Organic
Loading
Ib BOD/day
Ib MLSS
0.86
1.56
1.10
1.02
1.56
1.0
1.9
0.92
1.20
0.95
1.10
2nd-Stage
BOD Removal
Efficiency
percent
67.0
53.7
40.4
42.5
59.0
41.6
72.2
79.0
16.0
38.5
30.0
Ist-Stage
Volumetric BOD Removal
Loading Efficiency
Ib BOD/day
l,000cuft
182
197
160
128
188
205
375
154
196
158
186
Air Supplied
per Ib of
BOD Removed2
cu ft
4,600
4,600
4,600
4,600
4,600
4,600
4,100
4,100
4,100
4,100
4,100
percent
82
89
87
83
82
96
89
82
81
88
93
Aeration
Detention
Timel
hours
1.6
1.7
2.5
2.9
1.8
-
0.7
0.7
0.7
0.7
0.7
Air Supplied
per Ib of
BOD Removed
cuft
1,180
1,180
1,180
1,180
1,180
820
820
820
820
820
820
Overall
Secondary
BOD Removal
percent
94
95
93
93
93
95
96
96
81
93
95
Aeration
Detention
Time1
hours
1.6
1.7
2.5
2.9
1.8
0.7
0.7
0.7
0.7
0.7
0.7
Reference
28
29
Excluding sludge recycle.
Including that needed for ammonia oxidation.
-------
FIGURE 5-7
SCHEMATIC DIAGRAM OF MULTI-STAGE OXYGEN AERATION SYSTEM (31)
AERATION TANK COVERS
-SURFACE AERATOR
^MIXER DRIVE
OXYGEN
FEED GAS -*-
WASTEWATER
FEED *-Z
RECYCLE
S L II n R F *
3
-**-+
L_n W
-o
f
/
J
^^\,
->
L
-, / Jt EXHAUST
If | 1 1 =* R4C
Lx*^v
C. y^-^
i
i
I
I
I
1
C>0
\\
I
i
i
i
i
i
i
1
i
i
i
i
i
i
i
i
*
[1 BAFFLE
MIXED LIQUOR
=-» EFFLUENT TO
CLARIFIER
SUBMERGED PROPELLER (OPTIONAL)
5-18
-------
recommended to provide standby and peak load capacity. In the covered reactor design,
oxygen transfer and mixing are accomplished with surface aerators for tank depths up to 15
to 18 feet (as shown on Figure 5-7) or alternatively with submerged turbine-spargers and
recirculating gas compressors for deeper tanks.
Pure oxygen systems are normally designed to handle volumetric BOD loadings ranging from
150 to 225 Ib BOD/day/1,000 cu ft, but in some instances much higher loadings have been
reported (31)(32)(33)(34)(35)(36).
The F/M may vary from 0.07 to more than 1.0, but normal design values range from 0.5 to
0.8 (31)(32)(33). Table 5-8 summarizes operating data from plants utilizing pure oxygen.
Pure oxygen pilot plant studies have indicated that at a given F/M oxygen consumption
increases with higher COD/BOD ratios, as shown on Figure 5-8 (33). This increase in oxygen
requirements may be attributed to the biodegradation of part of the COD which is not
measured in the standard BOD test. Oxygen requirements for pure oxygen systems for
typical domestic wastewaters normally range from 0.6 to 0.7 Ib 02/lb COD removed in the
system (31).
Since the pure oxygen system operates at high MLSS concentrations, the clarifier will often
be subjected to increased solids loadings. Clarifier solids loadings for pure oxygen and other
systems are discussed in detail in Chapter 6. Clarifier underflow SS concentrations may
range from 1.0 to 3.5 percent, depending on the sludge composition and the clarifier sludge
detention time.
5.3 Activated Sludge Design Considerations
Initially, plant operators, through a trial and error procedure, developed the most efficient
operating criteria for conventional activated sludge plants as well as for the modifications of
the process. Out of this evolution, basic design criteria were developed. These criteria are
still in use today and, in many cases, are rigidly adhered to by regulatory agencies.
A limitation of these design criteria is that volumetric loading (Ib BOD applied/day/1,000 cu
ft) has been considered preferable, for design purposes, to organic or F/M loading (Ib BOD
applied/day/lb MLSS). Many of the modifications have shown that organic loading is an
important consideration; in fact, a higher volumetric loading has been achieved for
modifications of the conventional process at the same organic loading used in the
conventional process.
Basic parameters of interest in the design of an activated sludge process are:
1. BOD removal rates
2. Oxygen and air requirements
5-19
-------
TABLE 5-8
OPERATING DATA FROM PURE OXYGEN ACTIVATED SLUDGE PLANTS
BOD
Plant
Location
New York
Washington, D. C.
New York
Virginia
Indiana
Influent
Flow
mgd
1.33
1.29
1.38
1.19
1.36
1.41
1.64
0.07
0.10
0.10
20.8
17.7
15.1
20.6
25.3
30.0
35.4
11.0
4.4
Sludge
Recycle
percent
53
54
56
45
42
38
32
50
31.5
38
30
40
50
45
44
34
25
14
50
Secondary
Influent
mg/1
237
221
249
283
270
304
269
115
102
116
156
157
152
171
213
218
212
158
84
Secondary
Effluent
mg/1
22
18
19
19
9
11
15
19
12
14
9
21
17
17
22
21
23
14
10
Aeration
Tank
MLSS
mg/1
5,890
6,810
6,840
5,890
7,400
5,700
5,560
4,140
6,000
8,120
4,890
5,060
4,000
3,875
4,550
4,155
3,090
3,8202
5,150
Organic
Loading
Ib BOD/day
Ib MLSS
0.35
0.27
0.38
0.37
0.31
0.47
0.50
0.31
0.25
0.22
0.55
0.47
0.46
0.74
0.95
1.30
1.96
0.483
0.15
Volumetric
Loading
Ib BOD/day
1,000 cu ft
126
115
140
132
142
166
170
80
90
108
163
140
110
178
272
331
379
113
50
Aeration
Detention
Timel
hours
2.9
3.0
2.8
3.3
2.9
2.8
2.4
2.2
1.7
1.7
1.4
1.7
2.0
1.4
1.2
1.0
0.8
2.1
2.5
Secondary
BOD Removal
Efficiency
percent
91
92
93
92
97
97
94
84
88
88
94
87
89
90
90
90
89
91
88
Reference
37
38
26
39
40
1 Excluding sludge recycle.
2 Aeration tank MLVSS.
3 Ib BOD/day/lb MLVSS.
-------
FIGURE 5-8
OXYGEN CONSUMPTION FOR PURE OXYGEN SYSTEMS
RELATED TO ORGANIC LOADING (F/M) (33)
0.2
0,4 0,6 0.8
F/M, LB BOD/DAY/LB MLVSS
5-21
-------
3. Sludge production
4. Oxygen transfer rates in wastewater
5. Nutrient requirements
6. Separation and return of activated sludge.
5.3.1 BOD Removal Rates
Eckenfelder (41) has indicated that a linear arithmetic relationship exists between BOD
removal rate (mg BOD/hr/g VSS) and effluent BOD (mg/1), as shown on Figure 5-9 for
typical data from various complete-mix activated sludge plants. The variations in the BOD
removal relationship on Figure 5-9 are influenced by the presence of various proportions of
domestic and industrial wastes.
Using Figure 5-9, the detention time required to achieve a specific effluent BOD can be
obtained as follows:
. _ LaLe
Sar'
where:
t = Aeration tank detention time, hours
La = Influent BOD to aeration tank, mg/1
Le = Clarifier effluent BOD, mg/1
Sa = MLVSS, mg/1
r' = BOD removal rate, mg BOD/hr/g VSS
Weston (42) developed a log-log relationship between a BOD removal rate constant (r) and a
loading ratio (L0/S0); r and L0 are defined by the following equations:
r _ Lo~Le
where:
T i
L° " 1 + R
r = BOD removal rate constant,
Lo = BOD of wastewater after mixture of raw wastewater or
primary effluent with sludge recycle, mg/1
Le = Clarifier effluent BOD, mg/1
te = Aeration tank detention time, minutes (including recycle)
Lj = Raw wastewater or primary effluent BOD, mg/1
R = Sludge recycle as percent of influent flow
S0 = MLVSS, mg/1
5-22
-------
FIGURE 5-9
BOD REMOVAL CHARACTERISTICS
FOR VARIOUS COMPLETE MIX ACTIVATED SLUDGE PLANTS (41)
100
CO
oo
60
40
20
READILY REMOVABLE
ORGANICS
RESISTANT
ORGANICS
20 40 60 80
EFFLUENT BOD (rag/1)
00
5-23
-------
The aeration tank detention time is related to process efficiency (E, percent) and BOD
removal rate (r) by the following equation:
te =
100-E r
where:
E = ^5_ x 100
Lo
Operating data from over 20 plants with various activated sludge modifications were
analyzed using the Weston procedure to determine BOD removal rate constants. Operating
data for this analysis were taken from references (3)(4)(8)(12)(13)(15)(18)(21)(22)(23)
(28)(29)(37)(38)(43)(44)(45)(46)(47). The results are summarized on Figure 5-10. It
should be pointed out that the BOD removal rate curves represent only average kinetics with
no temperature correction applied, and the various loading ratios were determined using
MLSS, not MLVSS. For these reasons, these curves are not recommended for design
purposes, but are included merely to illustrate the relative kinetic rates of the modifications.
The presence of significant quantities of industrial wastes, which may have different removal
rate characteristics, would modify or displace the curves shown.
The BOD removal rate curve for the second stage of a two-stage process on Figure 5-10
represents data obtained from three two-stage biological treatment plants (28)(29)(45). Two
of these plants use activated sludge as the first stage, while the third plant uses trickling
filtration. The BOD removal rates for second-stage treatment are markedly lower than those
of conventional activated sludge because the organics remaining in the first-stage effluent are
more resistant to biological degradation than those entering a conventional plant.
5.3.2 Oxygen and Air Requirements
Table 5-9 contains ranges for oxygen and air required per Ib of BOD removed for the
activated sludge modifications previously discussed. These values represent overall process
requirements. The total amount of oxygen required will vary within the ranges shown
depending upon the F/M, increasing as F/M decreases. In diffused air systems, the air
requirements will vary depending on the oxygen transfer efficiency of the diffusers
employed. The designer must recognize that the various process modifications will require
different air and oxygen distribution patterns. Also, the values shown in the table do not
include allowances for nitrification or for other plant air requirements.
5-24
-------
FIGURE 5-10
RELATIONSHIP BETWEEN BOD REMOVAL RATE CONSTANTS
AND LOADING RATIOS FOR THE ACTIVATED SLUDGE MODIFICATIONS
003
LEGEND
CONVENTIONAL PROCESS
STEP AERATION PROCESS
O COMPLETE MIX PROCESS
4* PURE OXYGEN PROCESS
* CONTACT STABILIZATION
PROCESS (OVERALL r)
A CONTACT STABILIZATION
PROCESS (CONTACT r)
SECOND OF TWO-STAGE
PROCESS
02. ,03 .04 .05
LOAD RATIO Lo
. 10
5-25
-------
TABLE 5-9
OXYGEN AND AIR REQUIREMENTS FOR
ACTIVATED SLUDGE MODIFICATIONS
Process Ib Q2/lb BOD Removed Standard cu ft Air/lb BOD Removed
Conventional 0.8-1.1 800 - 1,500
Step Aeration 0.7 - 1.0 800 - 1,200
Contact Stabilization 0.7 - 1.0 800 - 1,200
Complete Mix 0.7 - 1.0 800 - 1,200
Modified Aeration 0.4 - 0.6 400 - 800
Pure Oxygen 0.8 - 1.4 -
5.3.3 Sludge Production
Normally, the activated sludge processes generate excess sludge in relation to the F/M
maintained in the system (48). For normal ranges of F/M (0.3 to 0.5 Ib BOD/day/lb
MLVSS), the quantity of excess sludge produced in air activated sludge systems varies
between 0.5 to 0.7 Ib VSS/lb BOD removed (21)(43)(49). The quantity of sludge increases
with increasing F/M, since at higher organic loadings less auto-oxidation occurs. The
quantity of sludge generated is also temperature dependent, decreasing with increasing
wastewater temperature. Data reported by various studies indicate that excess sludge
production in pure oxygen activated sludge systems may be slightly less than in air activated
sludge systems when operating at similar F/M's (31)(32)(49).
5.3.4 Oxygen Transfer Rates in Wastewater
Oxygen transfer rates in wastewater are affected by various physical and chemical
parameters, e.g., temperature, degree of mixing, liquid depth in the aeration tank, oxygen
composition of aerating gas, type of aeration device, operating DO, barometric pressure and
chemical characteristics of the wastewater. The major area often overlooked in the past by
design engineers has been the effect of an industrial waste on the overall oxygen transfer
rate of a system. Where the industrial waste makes up a large proportion of the total flow, it
is desirable to verify oxygen transfer rates in the laboratory for proper sizing of aeration
units.
Oxygen transfer capability for several aeration systems are indicated in Table 5-10. The
transfer capability for each system at standard conditions are approximate and may vary
somewhat for different makes and models. The effective transfer rates shown are for the
5-26
-------
specific field conditions listed in the table footnote, and should not be used for design
purposes. The relative effective transfer rates will vary considerably for different field
conditions, and must be determined for the design conditions at the specific installation.
TABLE 5-10
OXYGEN TRANSFER CAPABILITIES OF
VARIOUS AERATION SYSTEMS
Type of Aeration System
Diffused-Air, Fine Bubble
Diffused-Air, Coarse Bubble
Mechanical Surface Aeration, Vertical Shaft
Agitator-Sparger System
Pure Oxygen (50)
Mechanical Surface Aeration + Cryogenic Generation
Mechanical Surface Aeration + PSA Generation
Agitator-Sparger System + Cryogenic System
Standard
Transfer
Rate1
2.5
1.5
3.2
2.1
Effective
Transfer
Rate2
Ib O2/hp-hr Ib 02/hp-hr
1.4
0.9
1.8
1.2
2.6
1.9
2.2
1 Transfer Rate at standard conditions, i.e., tap water, 20 deg C, 760 mm barometric
pressure and initial DO = O mg/1.
2 Transfer rate at following specific field conditions
x = 0.85
6 = 0.9
t = 15 deg C
Altitude = 500 ft
Operating DO = 2 mg/1 for air aeration, 6 mg/1 for oxygenation
5.3.5 Nutrient Requirements
It is necessary that sufficient nitrogen and phosphorus be present in a wastewater so that
neither nutrient becomes the limiting factor in microbial growth reactions encountered in
the activated sludge process. Normally, supplemental nutrients are not required for
municipal wastewater treatment plants because adequate quantities are available in domestic
wastewaters to make organic carbon the limiting macronutrient. For optimum operation of
the activated sludge process, the minimum ratios of raw wastewater N:BOD and P:BOD are
3:60 and 1:60, respectively (51).
5-27
-------
5.3.6 Separation and Return of Activated Sludge
In the past, the importance of the final clarifier as an integral unit of the activated sludge
process has not been fully recognized. Improperly designed final clarifiers have resulted in
inefficient BOD and SS removals in many activated sludge plants.
Doubt has been raised as to the advisability of designing clarifier overflow rates solely on the
basis of average or nominal design flow. It is recommended that final clarifiers be designed
on the basis of maximum daily flow rates. This technique provides greater protection against
system solids washout, at the expense of a somewhat larger clarifier. Recent work by Dick
(52) also indicates that solids loading and thickening aspects must be considered in the
design of final clarifiers. Typical design parameters for final clarifiers are presented in Table
6-2.
Control of sludge recycle and wasting rates is the most important operational tool available
to the plant operator for intelligently managing the sludge inventory and maintaining
optimum loading conditions and sludge settling characteristics. Therefore, it is extremely
important to provide sufficient sludge recycle capacity to give the operator the required
operating flexibility to handle the highly variable and fluctuating waste loads characteristic
of many plants. Excessive detention of sludge within the final clarifier will result in
deterioration of the sludge.
Three techniques which have been used to control sludge recycle are:
1. Automatically varying the recycle flow to maintain a set relationship to influent
flow
2. Setting the recycle flow at a constant rate based upon the average daily flow
3. Controlling the recycle pumps with a sludge blanket sensor set to maintain a
predetermined blanket level in the final clarifier.
A firm sludge recycle capacity of at least 50 percent of average design flow is recommended
for the conventional and step aeration processes; at least 100 percent is recommended for
the contact stabilization and complete mix modifications. Firm capacity is defined as the
available pumping capacity with the largest pump out of service.
5.4 Pilot Studies
The use of pilot facilities for investigating the upgrading of existing activated sludge plants is
strongly indicated, in many cases, to ensure that optimum design parameters are selected.
There are two general types of piloting facilities available: batch or continuous-flow
systems. Batch studies are used primarily to evaluate treatability. Continuous-flow systems
5-28
-------
generally range in size from bench-scale to 10-gpm units and are used to generate data for
design, such as BOD removal rates, oxygen requirements and sludge production.
5.4.1 Batch Studies
Batch laboratory-scale units are subject to all of the inherent difficulties of biological
oxidation systems, with the added magnified complexities of large surface-to-volume ratios,
small quantities of sludge mass in the reactor and the undesirable factors associated with
slug feeding of wastewater. In spite of these inherent difficulties, batch studies have
attractive features in that they afford an economic and efficiently controlled method of
determining the biological treatability of a wastewater and developing fundamental
information concerning the applicability of various activated sludge modifications.
5.4.2 Continuous-Flow Studies
Use of the continuous-flow system is preferable to obtain design parameters since it
approximates the operation of an actual plant, permitting evaluation of the effects of
variations in wastewater loading or strength. Continuous-flow units must be used on
wastewaters which exhibit biostatic or exert toxic effects to permit development of an
acclimated biomass. However, most municipal wastewaters do not exhibit these properties
unless there is a significant discharge of untreated industrial wastes. A continuous-flow
bench-scale aeration unit is illustrated on Figure 5-11. The initial step in setting up such a
unit is to characterize the wastewater and to design the unit and select the proper loadings
accordingly. Frequently, the bench-scale unit is designed as an integral aerator-settler system
constructed of plexiglass with a volume of 1 to 10 liters. The aeration and settling
compartments are separated by a plexiglass baffle wall which may be adjusted to vary the
size of the opening between the two zones. Solids separated in the settling compartment
may be returned to the aeration compartment by sloping the bottom of the settling zone
toward the aeration compartment or preferably by the use of a small pump. Air is supplied
through a porous diffuser at the end of a tube. An effluent overflow tube is located at the
desired water level. The aeration unit may be fed either by a small metering pump, or by
gravity as illustrated on Figure 5-11. Once equilibrium conditions have been established
within the unit, the removal of BOD, nitrogen and phosphorus can be evaluated. SS removal
is determined from settling tests and the resulting sludge may be evaluated for
dewaterability by using a capillary suction device. An approximate determination of the
quantity of sludge produced per unit of BOD removed may also be made during a
bench-scale study.
A schematic of a medium-sized (0.1 to 1.0 gpm) continuous-flow pilot unit is shown on
Figure 5-12. Basically, the system consists of a wastewater feed tank equipped with a mixer
to blend the wastewater prior to feeding the aeration unit and to prevent solids deposition
in the feed tank. The wastewater from the feed tank and the recycle sludge from the final
clarifier are pumped to the aeration tank using peristaltic-type pumps. The difficulty of
5-29
-------
FIGURE 5-11
BENCH-SCALE AERATION UNIT
RATE CONTROLLER-
ADJUSTABLE BAFFLE
SETTLING
COMPARTMENT
WASTEWATER
CONTAINER
5-30
-------
FIGURE 5-12
SCHEMATIC OF A CONTINUOUS-FLOW PILOT UNIT
V ROTOMETER
WASTE«ATER
FEED TANK
MIXED LIQUOR
FEED PUMP
(PERISTALIC
TYPE)
POROUS
DIFFUSER
AERATION
TANK
SLUDGE RECYCLE
EFFLUENT
FINAL
CLARIFIER
SLUDGE PUMP
(PERISTALTIC TYPE)
EXCESS SLUDGE
5-31
-------
pumping extremely low flows makes it essential to provide a reasonably large
aerator-clarifier system. Air is normally supplied through porous diffusers controlled by the
use of rotameters. The wastewater, after treatment, flows by gravity to a separate clarifier.
In the small-scale clarifier, care must be taken to prevent solids deposition on the side walls.
The clarifier should have a scraper mechanism to aid in thickening and removal of the mixed
liquor solids. If 24-hour composite sampling of feed wastewater and clarifier effluent is
required, provision should be made to pump these streams into refrigerated sample bottles.
Two approaches may be applied for the acclimation and growth of a culture of
microorganisms for use in a pilot study. An available activated sludge culture may be
utilized as the source of microorganisms, with the normal feed to that system being
gradually replaced by the wastewater under investigation until satisfactory performance is
obtained. Alternatively, culture development can begin with a small quantity of seed
organisms and a wastewater feed diluted below the toxicity threshold (if toxicity exists). As
the biological mass develops, the toxicity threshold is redetermined and the wastewater
concentration is increased accordingly until the culture is capable of handling wastewater at
100 percent concentration. The latter technique is preferred because it provides the best
opportunity to observe the growth characteristics of the biological culture as well as
potential problems with acute or chronic toxicity.
When the culture is capable of functioning on the undiluted wastewater, data are collected
on the performance of the system, beginning with a low-feed rate and increasing the feed
rate until anticipated design loadings and performance are attained. For various F/M's, the
performance and characteristics of the system should be evaluated in terms of:
1. BOD, COD and SS removal
2. Oxygen consumption
3. Waste sludge production
4. Biomass characteristics (microscopic appearance and settling rates)
5. Physical nature of the effluent (turbidity, odor, color, etc.).
5.4.2.1 BOD Removal Rate Determinations
The data collected from continuous-flow units can be analyzed using either the Eckenf elder
or Weston procedures to define the appropriate BOD removal characteristics for design
conditions as previously discussed (41)(42).
5-32
-------
5.4.2.2 Oxygen Uptake Requirements
Reliable oxygen consumption data cannot be obtained from pilot plants because of the scale
factors involved. However, oxygen uptake studies can and should be conducted on the
mixed liquor to obtain energy and endogenous oxygen requirements. A schematic of an
oxygen uptake curve for a typical continuous-flow activated sludge pilot unit is shown on
Figure 5-13 (51). The slope of the line (m) represents the oxygen required for cell synthesis,
while the ordinate intercept (b) represents the oxygen required for endogenous respiration.
The net oxygen consumption is expressed by the following equation:
(\\> BOD removed A
62 - m I ~j~~I + b (lb VSS under aeration)
where:
O2 = lb oxygen uptake/day
m = energy oxygen, lb oxygen uptake/lb BOD removed
b = endogenous oxygen, lb oxygen uptake/day/lb VSS under aeration
5.4.2.3 Sludge Production
Sludge production in an activated sludge system is expressed as the net effect of two
processes as follows:
1. Synthesis of new organisms resulting from the assimilation of the organic material
removed
2. Reduction of the weight of organisms under aeration by the process of
endogenous respiration.
Figure 5-14 is a schematic representation of sludge production from a continuous-flow pilot
plant (51). The slope of the line (m') represents sludge synthesis, while the ordinate
intercept (b') represents the endogenous destruction of solids. The net sludge production is
expressed by the following equation:
VSS produced/day = m' (lb BOD removed/day) - b' (lb VSS under aeration)
where:
m' = sludge synthesis (lb VSS produced/lb BOD removed)
b' = endogenous destruction of sludge (lb VSS destroyed/day/lb VSS
under aeration)
5-33
-------
FIGURE 5-13
DETERMINATION OF OXYGEN UPTAKE REQUIREMENTS (51)
LB BOD REMOVED DAY LB MLVSS UNDER AREATION
ffl = ENERGY 02 (LB 02/LB BOO REMOVFn)
b = ENDOGENOUS 02 (LB 02/DAY,LB MLVSS UNDER AERATION)
5-34
-------
FIGURE 5-14
DETERMINATION OF SLUDGE PRODUCTION CHARACTERISTICS (51)
CO
CO
CO
CO
LB BOD REMOVED/DAY/LB MLVSS UNDER AERATION
m' = SLUDGE SYNTHESIS (LB VSS/LB BOD REMOVED)
b' = ENDOGENOUS DESTRUCTION OF SLUDGE (LB VSS/OAY/LB
MLVSS UNDER AERATION)
5-35
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Although the bench-scale and small pilot-scale techniques described above are extremely
useful in initial screening and treatability tests, extreme caution must be exercised in
extrapolating pilot-generated design data to full-scale plants. The pilot tests should be
conducted to approximate the most severe conditions expected for the full-scale plant.
5.5 Activated Sludge Upgrading Techniques and Design Bases
Upgrading activated sludge plants to relieve overloaded conditions, to improve organic
removal efficiency, to provide nitrification and to remove nutrients will be covered in the
following sections.
5.5.1 Upgrading to Relieve Overloaded Conditions
The following activated sludge modifications are examined as they apply to upgrading
existing activated sludge plants:
1. Step aeration
2. Contact stabilization
3. Complete mix
4. Pure oxygen.
The step aeration and contact stabilization processes are similar in that both modifications
can be incorporated into the upgraded design at a minimum capital investment. The
flexibility to operate in both modes is accomplished by sizing the influent step aeration
piping so that the entire flow may be introduced in the last bay of the aeration tank, thus
permitting operation as a contact stabilization process.
Before examining each individual upgrading procedure, several general statements can be
made. Operating data, BOD removal rate constants and volumetric loadings previously
discussed indicate that at least one of the activated sludge modifications may be applicable
for upgrading an overloaded activated sludge plant. These modifications may require
renovation of the air system to supply sufficient air to meet the requirements of the new
process because of higher volumetric loadings. The mechanical aerator and agitator-sparger
systems are illustrated on Figure 8-2. Even though mechanical aerators afford a high transfer
efficiency, their use in an existing basin may pose problems because the geometric
configurations required for their most efficient utilization may be quite different than the
existing basin configuration. This was found to be true for an upgrading investigation
performed for the City of Baltimore, Maryland (53). In an economic comparison, it was
found that the annual costs for a diffused versus a mechanical aeration system were
approximately equal because of the existing configuration of the plug flow basins. However,
it was recommended that mechanical aeration be considered for future aeration tank
expansion.
5-36
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Most existing conventional plants use either fine- or coarse-bubble diffused air systems. The
fine-bubble system is more efficient, but on the other hand represents a greater capital
investment and more costly maintenance problem than a coarse-bubble system. At
Milwaukee, Wisconsin, it was found that the type and arrangement of fine-bubble diffusers
had a significant effect on oxygen transfer efficiency. The studies showed that ceramic tubes
in both spiral and transverse patterns did not perform as efficiently as ridge and furrow
pattern ceramic plates located longitudinally. By converting to the ridge and furrow pattern,
the overall plant removal efficiency was substantially increased (54).
The agitator-sparger system has an operational advantage over the diffused air unit (coarse-
or fine-bubble) in that during low flows the air may be reduced but the mixing will be
maintained due to the action of the turbine agitator.
5.5.1.1 Upgrading a Conventional Activated Sludge Plant to Step Aeration
Step aeration has been used successfully as an upgrading technique in New York City;
Indianapolis, Indiana; and numerous other locations. The case history discussed below
illustrates the implementation of this technique of the Wards Island Wastewater Treatment
Plant in New York City (55)(56)(57).
Wards Island was built in 1937 to provide secondary treatment for a design flow of
180 mgd. By 1947, the average flow was in excess of 226 mgd, solids removal performance
had become erratic and air consumption was excessive. The problems were remedied and the
capacity was increased to 240 mgd by: (1) converting the conventional activated sludge
aeration tanks to the step aeration process and (2) modifying the final clarifier inlet-outlet
configurations.
The original design provided parallel treatment trains, with an aeration detention time of
7.0 hours and a final clarifier overflow rate of 700 gpd/sq ft at the design flow. The
modification of the aeration tank influent systems to provide step aeration, as shown on
Figure 5-15, increased the plant's organic loading capacity and greatly enhanced operational
flexibility. Using the step aeration process, 12 of the original aerators had sufficient capacity
for 240 mgd, providing a nominal aeration detention time of 4.0 hours. A study of plant
capacity carried out in 1965 concluded that if the existing preliminary tanks were replaced
and the four surplus aerators converted into final clarifiers, the plant could be rated for at
least 250 mgd.
Operating and performance data for the upgraded plant from 1953 are shown in Table 5-11
for one of the four treatment plant batteries (Battery C).
5-37
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FIGURE 5-15
UPGRADING A CONVENTIONAL ACTIVATED SLUDGE PLANT
TO STEP AERATION
PRIMARY
EFFLUENT
AERATION
TANK
FINAL
CLARIFIER
RETURN SLUDGE
EFFLUENT
EXCESS SLUDGE
ACTIVATED SLUDGE SYSTEM BEFORE UPGRADING
(1 OF 16 PARALLEL MODULES)
PRIMARY
EFFLUENT
AERATION
TANK
EFFLUENT
RETURN SLUDGE
EXCESS SLUDGE
ACTIVATED SLUDGE SYSTEM AFTER UPGRADING
(1 OF 16 PARALLEL MODULES)
5-38
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TABLE 5-11
OPERATING AND PERFORMANCE DATA FOR THE
WARDS ISLAND PLANT, NEW YORK CITY
Operating and Performance After
Description Upgrading (1953) (55) (58)
Flow, mgd 481
Raw Wastewater
BOD, mg/1 123
SS, mg/1 145
Primary Effluent
BOD, mg/1 99
SS, mg/1 89
Aeration Tanks
MLSS, mg/1 1,030
Air Rate, cu ft/gal 0.60
Volumetric Loading, Ib BOD/day/1,000 cu ft 46
Organic Loading, Ib BOD/day/lb MLSS 0.71
Detention Time, hr 3.32
Final Clarifiers
Overflow Rate, gpd/sq ft 900
Secondary Effluent
Overall BOD Removal, percent 95
Overall SS Removal, percent 95
BOD, mg/1 6
SS, mg/1 7
1 Battery C of 4 batteries.
2 Excluding sludge recycle.
5-39
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The costs for the plant modifications projected to an EPA cost index of 175 are as follows:
Final Clarifier Alterations $ 990,000
Aeration Tank Modifications 510,000
$1,500,000
5.5.1.2 Upgrading a Conventional Activated Sludge Plant to Contact
Stabilization
Contact stabilization has been successfully used as an upgrading technique in Austin, Texas;
York, Pennsylvania; and Bergen County, New Jersey. The case history discussed below
illustrates the implementation of this technique at Austin, Texas.
The first full scale application of the contact stabilization concept was developed by Ullrich
and Smith (59)(60) at the Austin, Texas Wastewater Treatment Facility. In 1954, the entire
plant was converted from the original 6.0 mgd conventional activated sludge operating mode
to the "Biosorption" process, later to be renamed contact stabilization. Earlier field studies
had indicated that implementation of contact stabilization could effectively control a
long-standing sludge bulking problem and significantly increase plant capacity. The cause of
the sludge bulking was undetermined, and could be effectively controlled in the
conventional operating mode only by bypassing a portion of the primary effluent.
Schematics of the original and upgraded facility are shown on Figure 5-16. Conversion of
the plant to contact stabilization was achieved through the splitting of the flow to two
separate parallel and independent plants. The two original aeration tanks were split between
the two plants, with additional clarification capacity provided. In the upgraded plant,
aeration tank A was coupled with the two existing final clarifiers and the original primary
clarifier converted to final clarification service. Aeration tank B was matched with two new
clarification units. Primary clarification was omitted in the upgraded contact stabilization
flow pattern. Table 5-12 compares the operation and performance of the facility before and
after upgrading. The data shown for the upgraded plant are for one-half of the treatment
facility (Plant A).
Overall removals averaged 94 percent for BOD and 92 percent for SS in January, 1956.
Sludge bulking problems were also eliminated. The data reveal that through conversion to
the contact stabilization process and the addition of further final clarification capacity, the
upgraded facility could adequately treat over twice the original average daily design flow.
5-40
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FIGURE 5-16
UPGRADING A CONVENTIONAL ACTIVATED SLUGE PLANT
TO CONTACT STABILIZATION
RAW
WASTEWATER
AERATION
TANK NO.
AERATION
TANK NO 2
COMBINED PRIMARY
AND WASTE SLUDGE
EFFLUENT
ACTIVATED SLUDGE SYSTEM
BEFORE UPGRADING
RETURN
SLUDGE
EXCESS
SLUDGE STABILIZATION
ZONE
SLUDGE STABILIZATION
ZONE
RAW WASTEWATER
(CONVERTED
EXISTING
PRIMARY
CLARIFIER)
EXCESS SLUDGE
ACTIVATED SLUDGE SYSTEM AFTER UPGRADING
5-41
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TABLE 5-12
OPERATING AND PERFORMANCE DATA FOR AUSTIN, TEXAS
Overloaded Operating Upgraded Operating
and Performance and Performance
Data1 Data2 (Plant A)
Flow, mgd 8.2 6.5
Raw Wastewater
BOD, mg/1 249 358
SS, mg/1 243 247
Primary Treatment
Primary Effluent BOD, mg/1 ,151
BOD Removal, percent 39
Primary Effluent SS, mg/1 98 -
SS Removal, percent 60
Aeration Tank
MLSS,mg/l 1,282 2,533
Sludge Recycle, percent 26.4 56
Air Requirement, cu ft air/lb
BOD removed 1,778 776
Volumetric Loading, Ib BOD/day/
1,000 cu ft 40.9 1383
Organic Loading, Ib BOD/day/lb MLSS 0.51 0.383
Detention Time in Aerator, minutes^1 331 59^
Detention Time in Stabilization
Zone, minutes^ 312
Secondary Clarifier
Overflow Rate, gpd/sq ft 774 410
Secondary Treatment
BOD Removal, percent 40 94
SS Removal, percent 70 92
Effluent BOD, mg/1 91 21
Effluent SS, mg/1 73 21
1 January - September, 1946.
2 January, 1956.
«* Including stabilization.
4 Excluding sludge recycle.
^ Contact zone.
" Based on return sludge flow.
5-42
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5.5.1.3 Upgrading a Contact Stabilization Plant to Complete Mix Activated
Sludge
Past experience with complete mix activated sludge on a large scale has been quite
successful, although somewhat limited. Complete mix plants have been installed at Grand
Island, Nebraska; Freeport, Illinois; South Tahoe, California; and Albany, Oregon.
McKinney (21) and Smith (44) have reported the usefulness of this process for upgrading an
overloaded activated sludge plant. When the complete mix process is to be considered as an
upgrading technique, the geometric configuration of the existing aeration basin sometimes
poses a major problem.
Coralville, Iowa (15) had a contact stabilization package plant which was providing
detention times of 2.6 and 6.5 hours, respectively, in the contact and stabilization zones,
based on the average influent flow of 867,000 gallons per day. The contact stabilization
package plant is shown in plan view on Figure 5-17.
As previously discussed in Section 5.2.3, a contact-zone detention time of this magnitude
may result in poor quality effluent because the sludge becomes partially stabilized in this
zone and exhibits poorer settling characteristics. After investigation, this was found to be
the case in Coralville. Operating and performance data from the plant before upgrading are
summarized in Table 5-13. The effluent BOD and SS concentrations were 26 and 24 mg/1,
respectively.
To improve the plant's performance, it was decided to modify the flow pattern as indicated
on Figure 5-17. The influent piping was modified so that the raw wastewater was evenly
distributed into what originally was the stabilization zone. No raw wastewater was
introduced into the former contact zone. Mixed liquor in the upgraded system proceeded
from the former stabilization zone through the former contact zone to the final clarifier.
The return sludge was introduced into the former stabilization zone at one point only.
Therefore, the upgrading resulted in a "modified" complete mix flow pattern, with an
overall detention time of 9.1 hours at an average influent flow of 867,000 gpd.
Operating and performance data for the upgraded plant are also included in Table 5-13. The
effluent BOD and SS concentrations were lowered to 13 and 6 mg/1, respectively, by the
upgrading procedure. The costs associated with this modification are primarily for piping
changes. No cost breakdown was available for the modification.
5.5.1.4 Upgrading Existing Treatment Plants to Oxygen-Activated Sludge
An existing activated sludge system may be upgraded by conversion to a pure oxygen
system. However, because of a number of distinct differences between the physical
5-43
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FIGURE 5-17
UPGRADING A CONTACT STABILIZATION PLANT TO
COMPLETE-MIX FLOW PATTERN ACTIVATED SLUDGE (15)
SLUDGE _ ^ ^-CONTACT ZONE
STABILIZATION ZONE
RETURN SLUDGE
EXCESS SLUDGE
WASTING LINE
BEFORE UPGRADING
CONTACT STABILIZATION FLOW PATTERN
RAW
WASTEWATER
EFFLUENT
AEROBIC DIGESTER
FINAL CLARIFIER
COMPLETE-MIX
AERATION TANK
RETURN
SLUDGE
EXCESS SLUDGE
WASTING LINE
AFTER UPGRADING
MODIFIED COMPLETE -MIX FLOW PATTERN
AERATION TANK
RAW
\ WASTEWATER
EFFLUENT
AEROBIC DIGESTER
FINAL CLARIFIER
5-44
-------
TABLE 5-13
OPERATING AND PERFORMANCE DATA
FOR CORALVILLE, IOWA
Parameter
Average Flow, mgd
Raw Waste water
BOD, mg/1
SS, mg/1
Aeration Tank
Sludge Recycle, percent
Contact Zone Volumetric Loading,
Ib BOD/day/1,000 cu ft
Contact Zone Organic Loading,
Ib BOD/day/U) MLVSS
Contact Zone MLSS, mg/1
Complete Mix Volumetric Loading,
Ib BOD/day/1,000 cu ft
Final Clarifier
Overflow Rate^, gpd/sq ft
Overall Plant Performance
BOD Removal, percent
SS Removal, percent
Effluent BOD, mg/1
Effluent SS, mg/1
Contact "Modified"
Stabilization 1 Complete Mi
0.867
0.867
135
150
60
78
0.4
3,500
135
150
60
750
81
84
26
24
224
750
90
96
13
6
1 Before Upgrading.
2 After Upgrading.
o
Based on average flow.
4 Based on total aeration volume.
5-45
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components of air and oxygen systems, special consideration must be given to the
following:
1. Tank foundations and walls must be checked structurally for increased loadings
due to the oxygen dissolution equipment and tank covers.
2. Baffling may be required to sectionalize the existing aeration tank for
compatibility with various pure oxygen dissolution configurations.
3. Existing air diffusion piping may have to be removed.
4. Protection must be provided against the potential explosion hazard of pure
oxygen or oxygen-enriched air.
5. Protection may have to be provided against potential accelerated corrosion due to
pure oxygen or oxygen-enriched air.
Aspects of using pure oxygen which may make it economically attractive include:
1. Significant increases in volumetric loading may be accommodated in existing
aeration tank structures.
2. The oxygen generation equipment may be placed outside and does not require a
protective enclosure.
3. Sludge production may be reduced.
4. Expensive renovation of the blower building is eliminated.
5. Liquid oxygen storage may be used to reduce peak energy requirements.
Full scale use of pure oxygen for upgrading municipal treatment plants handling extremely
large flows is in the startup stage in Detroit, Michigan, and is under consideration in New
York City. Oxygenation at Detroit has been utilized to expand a 300-mgd section of the
existing primary treatment plant to secondary treatment. In New York City, a 20-mgd
section of the Newtown Creek Wastewater Treatment Plant has been converted to oxygen
aeration to upgrade treatment efficiency. This conversion is presented as a case history
below (31).
The Newtown Creek Plant has 16 parallel bays, each of which has a design capacity of 20
mgd and consists of an aerated grit chamber, an aeration tank and a final clarifier. The
system is designed as a modified aeration process, a process that typically achieves 65 to
70 percent BOD removal when operated at design flow and MLSS concentration. Currently,
5-46
-------
the Newtown Creek facility is receiving only 55 to 60 percent (11 to 12 mgd/bay) of the
design flow. This temporary flow condition has allowed operation of the modified aeration
system in a more conventional mode with a nominal aeration detention time of
approximately 2.5 hours and an MLSS concentration of 2,000 mg/1, as shown in Table 5-14.
When design flow is reached within the next one to two years, operation will have to revert
to the modified aeration mode because of air supply limitations. At that time, the City will
be confronted with a severe upgrading problem in a land-locked neighborhood.
TABLE 5-14
OPERATING AND PERFORMANCE DATA FOR
NEWTOWN CREEK PLANT,
BROOKLYN, NEW YORK
Oxygen2
Description System (26) System (31)
Flow, mgd 10.9 20.6
Raw Wastewater
BOD, mg/1 202 171
SS, mg/1 184 159
Aeration Tank
MLSS, mg/1 2,000 3,875
Sludge Recycle, percent 20 45
Oxygen Requirements, Ib 02/lb BOD removed 16.1 3 1.02
Volumetric Loading, Ib BOD/day/1,000 cu ft 106 178
Organic Loading, Ib BOD/day/lb MLSS 0.92 0.74
Detention Time in Aerator, minutes^ 158 86
Secondary Clarifier
Overflow Rate, gpd/sq ft 500 936
Secondary Treatment
BOD Removal, percent 88 90
SS Removal, percent 80 89
Effluent BOD, mg/1 24 17
Effluent SS, mg/1 37 18
1 October 1, 1972 through November 30, 1972.
2 Phase 4 of study, April 8, 1973 through June 2, 1973.
3 Excludes air diffuser oxygen transfer efficiency.
4 Excluding sludge recycle.
5-47
-------
Pure oxygen is believed to be a good candidate for achieving the required 90 percent BOD
removal within the confines of the existing aeration and final clarification tankage at design
flow. To evaluate this possibility, one bay of the Newtown Creek Plant was converted to a
staged, covered reactor oxygen system, as shown on Figure 5-18. This test system has been
operated continuously since May, 1972.
From September 17, 1972 to September 1, 1973, seven phases of a multi-faceted
experimental program were conducted. The oxygenation system was evaluated under
different loading and diurnal flow patterns in each phase. Operating and performance data
for Phase 4 are summarized in Table 5-14. In this phase, the average flow approximated the
design flow of the bay, with diurnal fluctuation matching the typical diurnal pattern
experienced at the plant. Throughout the seven phases, the pure oxygen system achieved
overall BOD removals of at least 87 percent. In Phase 4, the BOD removal equalled the
required rate of 90 percent.
A second case history wherein an existing plant was converted to pure oxygen is described
below for Fairfax County, Virginia. In May, 1970, Fairfax County was required to upgrade
its Westgate Treatment Plant to prevent the imposition of a building moratorium in the area
served by this plant (61). The original plant was designed in 1954 as an 8-mgd capacity
facility consisting of two rectangular tanks baffled to form three compartments, as shown
on Figure 5-19. The plant was designed to achieve 50 percent BOD removal. By 1970,
average flows had increased to 12 mgd and BOD removal was reduced at times to as low as
35 percent.
The Westgate Plant was scheduled to be abandoned by 1975 when its service area is to be
incorporated into a regional system tributary to an advanced waste treatment facility.
Hence, a minimum cost interim upgrading plan was sought to meet the imposed
requirements of 80 percent BOD removal. It was decided to upgrade the facility with pure
oxygen-activated sludge.
Portions of the existing rectangular tanks were converted to oxygenation reactors. The
initial clarification zone was maintained, and the remainder of the tanks was divided
longitudinally to form two reactors each, for a total of four. Each reactor was then baffled
to form four stages. The first three stages were covered and the fourth left open to the
atmosphere. Surface aerators were provided in each stage, but oxygenation occurred in the
first three stages only. Two circular final clarifiers were constructed. The upgraded facility is
shown on Figure 5-19. Since this was only a short-term upgrading, the pure oxygen supply
system selected was simply an oxygen storage station in which trucked-in liquid oxygen is
stored, vaporized and fed to the reactors. The upgraded facility became operational in
October, 1971.
5-48
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FIGURE 5-18
UPGRADING A MODIFIED AERATION ACTIVATED SLUDGE SYSTEM
TO OXYGEN-ACTIVATED SLUDGE
PUMPS
RAW
»ASTE»ATER-
r-GAS RECIRCULATING
\COMPRESSORS
SUBMERGED PROPELLER-
s
1
)
>
-*
\ /SPARGER ASSEMBLY
ODD D D /
o
o
o
o
o
o
/
0
o
fe
fc
Ju
JO
1 In
-200'-
J I
-400'
EFFLUENT
PLAN VIEW
RAW
»ASfE«IATER »
GRIT
CHAMBER
Ml XER DRIVE FOR
V PROPELLER-SPARGER
\ ASSEMBLY
Yi n
n
I
FOUR-STAGE OXYGEN REACTOR
SKD = ,5
"Z
RETURN SLUDGE
EXCESS SLUDGE
ELEVATION
TREATMENT SYSTEM AFTER UPGRADING TO OXYGENATION (31)
5-49
-------
FIGURE 5-19
UPGRADING A PRIMARY TREATMENT PLANT
TO PURE OXYGEN ACTIVATED SLUDGE TREATMENT
CLAR
RAW
WASTEWATER >
IFIER
s AERATION* CLARIFIERS-,
/ BASINS-y ~7
i
i
RAW
WASTEWATER
EFFLUENT
*WITHOUT SLUDGE RECYCLE
BEFORE UPGRADING
PRIMARY
CLARIFIERS
BIOLOGICAL REACTOR
3 STAGES OXYGENATION
1 STAGE AIR AERATION
7
EFFLUENT
AFTER UPGRADING EXCESS
SLUDGE
5-50
-------
Operating and performance data for the original facility and the upgraded plant are listed in
Table 5-15. BOD removals greater than 90 percent are presently being achieved. The cost of
upgrading this facility, adjusted to an EPA cost index of 175, was $907,000.
TABLE 5-15
OPERATING AND PERFORMANCE DATA FOR
FAIRFAX COUNTY, VIRGINIA
Before Upgrading After Upgrading
Description (1969) (62) (May, 1972) (63)
Average Flow, mgd 11.3 11.0
Influent BOD, mg/1 245 158
Influent SS, mg/1 - 148
Aeration Tank
MLVSS, mg/1 _ 3,822
Sludge Recycle, percent 13.7
Oxygen Requirements, Ib O2/lb BOD Removed - 0.8-1.052
Volumetric Loading, Ib BOD/day/1,000 cu ft - 113
Organic Loading, Ib BOD/day/lb MLVSS - 0.48
Detention Time in Aerator, minutesl 126
Secondary Clarifier
Overflow Rate, gpd/sq ft 509
Solids Loading, Ib/sq ft/day 23.3
Overall Treatment
BOD Removal, percent 49 04
SS Removal, percent _ 87
Effluent BOD, mg/1 _ 14
Effluent SS, mg/1 _ 19
* Does not include sludge recycle.
The higher uptake rate occurs in warm weather operation due to the higher activity of the
biomass.
5-51
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5.5.2 Upgrading to Improve Organic Removal Efficiency
Upgrading techniques previously discussed relate to the ability of existing activated sludge
facilities to handle increased loads and/or overcome operational problems by providing
modifications to meet existing effluent standards. However, there may be a need to meet
higher effluent standards even though the existing facilities are not overloaded and are
performing in accordance with design expectations. Table 5-16 contains suggested
alternatives for improving effluent quality under these conditions, along with a range of
anticipated improvement in performance for each alternative.
It should be emphasized that, in cases where new unit processes are constructed downstream
of existing activated sludge facilities, the improvement in over ah1 organic removal will be a
direct function of the BOD removal achieved in the "add-on" process. However, where new
unit processes are installed upstream of existing activated sludge units, e.g., the use of a
roughing filter, the overall BOD removal may not be increased in direct proportion to the
amount achieved by the "add-on" process.
A detailed discussion on polishing lagoons, microscreens, filters, activated carbon and
clarifier modifications appears in subsequent chapters. The applicability of these alternatives
to individual cases should be evaluated in detail prior to the implementation of a particular
upgrading procedure.
5.5.3 Upgrading for Nutrient Control and Removal
With the trend toward more stringent effluent standards, the upgrading of activated sludge
plants to achieve phosphorus removal and nitrification and/or nitrogen removal is becoming
more frequent. Some of the techniques that may be utilized for this purpose are discussed
below. A case history for upgrading an existing plant for combined phosphorus and nitrogen
removal is presented in Chapter 13.
5.5.3.1 Phosphorus Removal
One of the most commonly used methods to accomplish phosphorus removal in an activated
sludge system is to precipitate the phosphorus by addition of a metallic salt to the
secondary system. Metallic salts that have been used include aluminum sulfate, sodium
aluminate, ferric chloride, ferric sulfate and ferrous sulfate. The selected salt may be applied
to the primary effluent as it enters the aerator, directly to the mixed liquor at a
predetermined point in the aerator or to the mixed liquor channel leading to the final
clarifier. The optimum addition point should be selected through a combination of
laboratory jar testing and comprehensive field testing. A flexible chemical addition design
permitting addition at several points in the secondary system flow pattern is highly
desirable. When sufficient metallic salt dosages are utilized, phosphorus removals of 80 to
5-52
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TABLE 5-16
UPGRADING TECHNIQUES FOR IMPROVEMENT OF ACTIVATED
SLUDGE TREATMENT PLANT EFFICIENCY
Addition Preceding
Existing Process
Roughing Trickling Filter
(Rock or Synthetic Media)
Chemical Addition
to Primary Clarifier
Existing Process
Activated Sludge
01
to
Addition Following
Existing Process
2nd-Stage Activated Sludgel
Polishing Lagoon
Multimedia Filters
Microscreens
Activated Carbon
Incremental BOD
Removal Across
the Added or
Modified Process
percent
20-40
30-50
30-70
30-60
50-80
30-80
60-80
1 A consideration if year-round nitrification is required.
-------
90 percent may be obtained. Case histories illustrating the implementation of phosphorus
removal in activated sludge plants may be found in the Process Design Manual for
Phosphorus Removal (64).
The addition of metallic salts will increase the solids loading on the final clarifiers, and
decrease the percentage of volatile solids in the clarifier underflow. If the benefits of
chemical addition are to be fully realized, the existing clarifiers must have adequate capacity
to handle the higher solids loadings. Also, the existing sludge recycle and processing systems
must be capable, respectively, of returning sufficient sludge to the aeration system to
maintain proper biological activity in that unit, and to handle the increase in the total
quantity of waste sludge generated by the process. If adequate existing capacity is lacking in
any of the above, additional construction may be necessary prior to implementation of
phosphorus removal.
5.5.3.2 Nitrification
If year-round nitrification is a design criteria, the information presented in Table 5-17 may
be useful in the preliminary sizing of second-stage process units (65). It should be stressed
that the detention time required to achieve nitrification is strongly dependent upon the
wastewater temperature and the MLVSS concentration maintained in the system.
TABLE 5-17
DESIGN GUIDELINES FOR NITRIFICATION
Description Design Parameter
Nitrification Reactor
Optimum pH range 7.6 to 7.8
Maximum influent BOD, mg/1 40 to 50
Tank configuration plug flow
MLVSS, mg/1 1,000 to 2,000
DO at average loading, mg/1 3.0
Minimum DO at peak loads, mg/1 1.0
Sludge recirculation, percent 50 to 100
Detention time based on average flow, hr 2 to 6
Oxygen requirements (stoichiometric), Ib 02/lb NHgN 4.6
Clarifier
Maximum allowable overflow rate, gpd/sq ft 1,000
5-54
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Nitrification reactors should be designed to ensure that plug flow will exist throughout the
reactor. This is because the rate of oxidation of ammonia nitrogen is basically linear,
requiring full utilization of the available detention time to complete the oxidation to the
higher valence form of nitrate nitrogen. Plug flow is probably best ensured by providing a
series of compartments within the reactor. Compartmentation has the further advantage
that reactors or portions of reactors may be taken out of service when ambient temperatures
are such that the full reactor design capacity is not required to accomplish nitrification.
Air requirements for completing the oxidation process can vary widely depending on pH,
wastewater temperature, diurnal flow variations and chemical characteristics of the influent
to the reactors (66). These changing oxygen requirements can be met with properly
monitored and controlled diffused air or mechanical aeration systems, or a combination of
the two.
If nitrification is not required during the colder months, a two-stage system may not be
necessary and nitrification may be accomplished within the existing single-stage activated
sludge system. The degree of nitrification that can be attained in a single-stage activated
sludge system depends on the SRT, mixed liquor DO concentration and wastewater
temperature (67). The SRT or rate of wasting sludge is of primary importance. If sludge is
wasted at too high a rate, the nitrifying bacteria will be washed from the system. Generally,
nitrification begins at an SRT of about five days, but does not become appreciable until the
SRT reaches about 15 days, depending upon the temperature. The existing single-stage air
supply system must be sufficient to provide the additional oxygen needed to oxidize the
ammonia nitrogen.
5.5.3.3 Nitrogen Removal
Denitrification may be added to the nitrification process as the second stage of the nitrogen
removal sequence. In this stage, nitrate is reduced to carbon dioxide, water and nitrogen gas
following addition of methanol to provide the carbon source. Other methods of removing
nitrogen from wastewaters include ammonia stripping, chlorination and ion exchange.
5.6 References
1. Metcalf & Eddy, Inc., Wastewater Engineering. McGraw-Hill Book Company, New
York City (1972).
2. Sawyer, C.N., Activated Sludge Modifications. Journal Water Pollution Control
Federation, 32, No. 3, p. 232 (1960).
3. Haseltine, T.R., A Rational Approach to the Design of Activated Sludge Plants.
Included in Biological Treatment and Industrial Wastes, edited by McCabe, J., and
Eckenfelder, W.W., Jr., Reinhold Publishing Company, New York City (1956).
5-55
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4. Phosphate Study at the Baltimore Back River Wastewater Treatment Plant. U. S. EPA,
Program Number 17010 DFV (September, 1970).
5. Operating Reports. Hyperion Treatment Plant, Sewage Treatment Division, Bureau of
Sanitation, City of Los Angeles, California (September, 1972).
6. Summary of Operation. Sewage Treatment Plant, Jeanette, Pennsylvania.
7. The Metropolitan Sanitary District of Greater Chicago, Calumet Treatment Works
Operating Data (1972).
8. Torpey, W., and Ghasick, A.H., Principles of Activated Sludge Operation. Included in
Biological Treatment of Sewage and Industrial Wastes, edited by McCabe, J., and
Eckenfelder, W.W., Jr., Reinhold Publishing Company, New York City (1956).
9. Allegheny County Sanitary Authority, Pittsburgh, Pennsylvania, Pittsburgh Sewage
Treatment Plant Operating Data (February, 1974).
10. The Metropolitan District, Hartford, Connecticut, Hartford Water Pollution Control
Plant Operating Data (September, 1973).
11. Metropolitan Department of Works, Toronto, Ontario, Canada, Main Treatment Plant
Operating Data (August, 1973).
12. Ullrich, A., and Smith, M., Operation Experience with Activated Sludge - Biosorption
at Austin, Texas. Sewage and Industrial Wastes, 29, No. 4, p. 400 (1957).
13. Grich, E., Operating Experience with Activated Sludge Reaction. Journal Water
Pollution Control Federation, 33, No. 8, p. 856 (1961).
14. Recommended Standards for Sewage Works. Great Lakes-Upper Mississippi River b
Board of State Sanitary Engineers (1971).
15. Dague, R.R., Elbert, G.F., and Rockwell, M.D., Contact Stabilization: Theory,
Practice, Operational Problems and Plant Modifications. Presented at the 43rd Annual
Conference of the Water Pollution Control Federation, Boston, Massachusetts
(October, 1970).
16. McKinney, R., Research and Current Developments in the Activated Sludge Process.
Journal Water Pollution Control Federation, 37, No. 12, p. 1696 (1965).
17. Full Scale Parallel Activated Sludge Process Evaluation. Freeport, Illinois Water and
Sewer Commission, U. S. EPA Project No. 17950 ENM (January, 1972).
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18. Private communication with C.L. Swanson, Sanitary Engineer, U. S. EPA, Cincinnati,
Ohio (November 5, 1970).
19. Robinson, J.H., New Regional Plant Built for the Future. Water and Wastes
Engineering, 10, No. 8, pp. 35-37 (August, 1973).
20. Private communication with Dr. Ross McKinney, University of Kansas, Lawrence,
Kansas (March 28, 1974).
21. McKinney, R., et al, Evaluation of a Complete Mixing Activated Sludge Plant. Journal
Water Pollution Control Federation, 42, No. 5, p. 737 (1970).
22. Hammer, M., and Tilsworth, T., Field Evaluation of a High Rate Activated Sludge
System. Water and Sewage Works, 115, No. 6, p. 261 (1968).
23. Private communication with M.E. Bolding, Water Reclamation Research Center, Dallas,
Texas (January, 1971).
24. Plant Performance Summary. Bureau of Wastewater Treatment, Government of the
District of Columbia (1971, 1972).
25. Schwinn, D.E., Design Features of the District of Columbia's Water Pollution Control
Plant. Presented at the Sanitary Engineering Specialty Conference, Sanitary
Engineering Division, ASCE, Rochester, New York (June, 1972).
26. Operating Data and Report. Department of Water Resources, New York City (May
1972 - January 1973).
27. Operating Reports. Department of Water and Sewers, City of Miami, Florida.
28. Private communication with Department of Civil Engineering, Pennsylvania State
University, University Park, Pennsylvania (January, 1968).
29. Simpson, R.W., Activated Sludge Modification. Water and Sewage Works, 106, No. 10,
p. 421 (1959).
30. Earth, E.F., Brenner, R.C., and Lewis, R.F., Chemical - Biological Control of Nitrogen
and Phosphorus in Wastewater Effluent. Journal Water Pollution Control Federation,
40, No. 12, p. 2,040 (1968).
31. Brenner, R.C., EPA Experiences in Oxygen-Activated Sludge. Prepared for U. S. EPA
Technology Transfer Design Seminar Program (October, 1973).
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32. Stamberg, J.B., Bishop, D.F., Hais, A.B., and Bennett, S.M., System Alternatives in
Oxygen Activated Sludge. Presented at the 45th Annual Conference of the Water
Pollution Control Federation, Atlanta, Georgia (October, 1972).
33. Operating Experience and Design Criteria for UNOX Wastewater Treatment Systems.
Union Carbide Corporation, Linde Division, Tonawanda, New York. Prepared for U. S.
EPA Technology Transfer Design Seminar Program (1972).
34. Report on Wastewater Treatment Pilot Plant Study for the City of Eculid, Ohio.
Havens and Emerson, Cleveland, Ohio (April 20, 1971).
35. Continued Evaluation of Oxygen Use in Conventional Activated Sludge Processing.
U. S. EPA Project No. 17050 DNW, Contract No. 14-12-867 (February, 1972).
36. UNOX Design Information for Contract Documents. Metcalf & Eddy, Inc., Boston,
Massachusetts. Prepared for U. S. EPA Technology Transfer Design Seminar Program
(1972).
37. Albertsson, J., McWhirter, J.R., Robinson, E.K., and Vahldieck, N.P., Investigation of
the Use of High Purity Oxygen Aeration in the Conventional Activated Sludge Process.
Federal Water Quality Administration, Program No. 17050 DNW, Contract
No. 14-14-465 (May, 1970).
38. Private communication with D.F. Bishop, Chief, U. S. EPA - Washington, D.C. Pilot
Plant, Washington, D.C. (January 10-11, 1971).
39. Air Products and Chemicals, Inc., OASES Wastewater Treatment Plant at the Fairfax
County, Virginia Westgate Treatment Plant, Release No. 2 (September 1, 1972).
40. Henry B. Steeg & Associates, Inc., Town of Speedway, Indiana Wastewater Treatment
Plant Design Data and Other Pertinent Information.
41. Eckenfelder, W.W., Jr. Theory of Design. Included in The Activated Sludge Process in
Sewage Treatment Theory and Application. Presented at a Seminar at the University of
Michigan, Ann Arbor, Michigan (February, 1966).
42. Weston, R.F., Fundamentals of Aerobic Biological Treatment of Wastewater. Public
Works, 94, No. 11, p. 74 (1963).
43. Torpey, W., Practical Results of Step Aeration. Sewage Works Journal, 20, No. 5,
p. 781 (1948).
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44. Smith, H., Homogeneous Activated Sludge-Three Parts. Water and Wastes Engineering,
4, No. 7, 8, 10, pp. 46-50, 56-63, 40-53 (1967).
45. Private communication with Leonard Waller, Plant Superintendent, South River Water
Pollution Control Plant, Atlanta, Georgia (January 27, 1971).
46. Boon, A.G., The Role of Contact Stabilization in the Treatment of Industrial Waste
and Sewage. Journal of Effluent and Water Treatment, 9, No. 6, p. 319 (1969).
47. Jackson, R., Bradney, L., and Bragstad, R.E., Short Term Aeration Solves Activated
Shidge Expansion Problems at Sioux Falls. Journal Water Pollution Control
Federation, 37, No. 2, p. 255 (1965).
48. Lesperance, T.W., A Generalized Approach to Activated Sludge. Reprinted from Water
and Wastes Engineering by Reuben H. Donnelly Corporation, New York, N.Y.
49. Union Carbide Unox System Wastewater Treatment. Union Carbide Corporation,
Linde Division, Tonawanda, New York (1970).
50. Private communication with Ted Zander, Linde Division, Union Carbide Corp.,
Tonawanda, New York (June 3, 1974).
51. Eckenfelder, W.W., Industrial Water Pollution Control. McGraw-Hill Book Company,
New York City (1966).
52. Dick, R., Role of Activated Sludge Final Settling Tanks. Journal of the Sanitary
Engineering Division, ASCE, 96, No. 2, p. 423 (1970).
53. Letter Report to the City of Baltimore, Maryland, Roy F. West on, Inc., West Chester,
Pennsylvania (July 14, 1970).
54. Leary, R.D., et al, Effect of Oxygen-Transfer Capabilities on Wastewater Treatment
Plant Performance. Journal Water Pollution Control Federation, 40, No. 7, p. 1,298
(1968).
55. Chasick, A.H., Activated Aeration at the Wards Island Sewage Treatment Plant. Journal
Sewage and Industrial Wastes, 26, No. 9, pp. 1059-1068 (September, 1954).
56. Gould, R.H., Wards Island Plant Capacity Increased by Structural Changes. Journal
Sewage and Industrial Wastes 22, No. 8, pp. 997-1003 (August, 1950).
57. Upgrading Existing Wastewater Treatment Plants Case Histories. U. S. EPA
Technology Transfer Design Seminar Publication, pp. 16-18 (August, 1973).
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58. Wards Island STP Step Aeration Operation Results, N.Y.C. Department of Public
Works, Division of Engineering (1953-1955).
59. Ullrich, A.H., and Smith, M.W., "The Biosorption of Process and Waste Treatment,"
Sewage and Industrial Wastes, Vol. 23, No. 10, pp. 1242-1253 (October 1951).
60. Ullrich, A.H. and Smith, M.W., Operation Experience with Activated
Sludge-Biosorption at Austin, Texas. Journal Sewage and Industrial Wastes, 29, No. 4,
pp. 400-413 (April, 1957).
61. Robson, C.M., Block, C.S., Nickerson, G.L., and Klinger, R.C., Operational Experience
of a Commercial Oxygen Activated Sludge Plant. Presented before the 45th Annual
Conference, Water Pollution Control Federation, Atlanta, Georgia (October 12, 1972).
62. Nickerson, G.L. and Van Atten, J.L., Westgate-A Study in Plant Modification.
Presented before the Chesapeake Water Pollution Control Association and Water and
Waste Operators Association of Maryland, Delaware and District of Columbia (June 8,
1973).
63. Westgate Plant Operating Data (April 29 through May 31, 1972).
64. Process Design Manual for Phosphorus Removal. U. S. EPA Office of Technology
Transfer, Washington, D.C. (revised 1974).
65. Sawyer, C.N., Design of Nitrification and Denitrification Facilities. Presented at a
Symposium on Design of Wastewater Treatment Facilities. Presented by U. S. EPA,
Cleveland, Ohio (April 22-23, 1971).
66. Nitrification and Denitrification Facilities. Prepared for U. S. EPA Technology
Transfer Program (1972).
67. Jenkins, D. and Garrison, W.E., Control of Activated Sludge by Mean Cell Residence
Time. Journal Water Pollution Control Federation, 40, No. 11, p. 1905 (1968).
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CHAPTER 6
CLARIFICATION AND CHEMICAL TREATMENT
6.1 Advantages of Upgrading Clarifiers
6.1.1 General
One of the more important considerations in any upgrading situation is the utilization of
overloaded clarification units. In many cases, it is possible to achieve substantial
improvements in their performance by the use of appropriate upgrading techniques.
By improving the performance of existing clarifiers, the requirement for new facilities can
be minimized. In a typical activated sludge plant, it can be expected that with an increased
capture of 1.0 mg/1 of SS, a concurrent reduction of about 0.6 mg/1 of BOD will be attained
in the final effluent. Consequently, the required capacity of new units providing either
physical or biological treatment may be reduced.
Clarifier performance can be improved in two ways. The first is to improve the physical
characteristics of the tanks to achieve more efficient settling and compaction. The second
approach is to introduce chemicals into the wastewater to improve the settling
characteristics of the SS.
6.1.2 Primary Clarifiers
The upgrading of primary clarifiers has the following advantages:
1. Improved BOD and SS removals at the same flow or similar BOD and SS removals
at somewhat higher flows.
2. An increase in quantity of primary sludge produced which can be more readily
thickened and dewatered than secondary sludge.
3. A decrease in quantity of secondary sludge produced.
4. A decrease in organic loading to secondary treatment process units.
Primary clarifier performance significantly influences the extent of secondary treatment
required and, in most cases, affects the overall effluent quality of existing treatment plants.
Also, since clarification is the most economical way to remove suspended and colloidal
pollutants, every effort should be made to improve the primary clarification process before
additional primary or secondary facilities are considered.
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6.1.3 Secondary Clarifiers
The performance of secondary wastewater treatment systems is determined by comparing
the quality of the overflow from secondary clarifiers to that of the incoming wastewater.
The biological treatment unit converts a portion of the soluble and insoluble organic
pollutants to suspended organic solids (biological). Unless these organic solids are effectively
removed in the secondary clarifiers, the treatment process cannot be considered a successful
operation. In fact, of all the process design variables which can affect overall plant
performance, those which are selected for secondary clarification are among the most
critical.
Effective secondary clarification also will allow more efficient disinfection, reduce the
frequency of cleaning chlorine contact tanks, and provide an aesthetically pleasing effluent.
6.2 Process Design of Clarifiers
6.2.1 Primary Clarifiers
Primary clarifiers are designed mainly on the basis of hydraulic overflow rates. In practice,
the designer will select appropriate overflow rates for the average and peak flows being
considered. Required surface areas are then calculated for both conditions and the larger
area is used. In many cases, especially in combined collection systems, the peak flow
conditions will govern the design. It should be recognized, however, that clarifier efficiency
at peak flows is a function of both magnitude and duration. Therefore, it is of extreme
importance that the designer analyze past flow data and the characteristics of the collection
system in order to avoid a design based on a short duration peak which may have little
effect on clarifier efficiency.
After selection of an overflow rate, the detention time in the clarifiers will be set by the
depth. Generally, SS in raw wastewaters are susceptible to some degree of flocculation.
Adequate detention time should be provided to allow this flocculation to occur and in turn
increase the clarifier efficiency. Normally, the depth should be set to provide a detention
time of 90 to 150 minutes.
For rectangular tank design, the longitudinal velocity of flow must also be considered.
Excessive velocities can result in scouring of settleable solids with subsequent losses to the
effluent. Peak velocities greater than four to five fpm should be avoided. Tank length to
width ratios of less than 3:1 should be avoided to prevent short-circuiting.
Typical design parameters for primary clarifiers are presented in Table 6-1. These parameters
are based on municipal wastewaters containing pollutants in the concentrations normally
encountered. For wastewaters containing a high percentage of industrial wastes, it may be
advantageous to initiate a testing program to determine design criteria. Based on the values
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presented in Table 6-1, it can be anticipated that a well-designed primary clarifier will be
capable of removing 30 to 35 percent of the BOD and 50 to 60 percent of the SS from a
domestic wastewater.
TABLE 6-1
TYPICAL DESIGN PARAMETERS FOR PRIMARY CLARIFIERS
Overflow Rate
Type of Treatment Average Peak
gpd/sq ft
Primary Settling Followed
by Secondary Treatment 800-1,200 2,000-3,000 10-12
Primary Settling with Waste
Activated Sludge Return 600-800 1,200-1,500 12-15
6.2.2 Secondary Clarifiers
The function of secondary clarifiers varies with the method of biological treatment utilized.
Clarifiers following trickling filters must effectively separate biological solids sloughed from
the filter media. Clarifiers in an activated sludge system serve a dual purpose. In addition to
providing a clarified effluent, they must also provide a concentrated source of return sludge
for process control. Adequate area and depth must be provided to allow this compaction to
occur while avoiding rejection of solids into the tank effluent. Secondary clarifiers in
activated sludge systems are also sensitive to sudden changes in flow rates. Therefore, the
use of multispeed pumps for in-plant wastewater lift stations is strongly recommended
where flow equalization is not provided.
The design of clarifiers following trickling filters is based on hydraulic overflow rates similar
to the method described for primary clarifiers. Design overflow rates must include
recirculated flow where clarified secondary effluent is used for recirculation. Because the
influent SS concentrations are low, tank solids loadings need not be considered. Typical
design parameters for clarifiers following trickling filters are presented in Table 6-2.
Clarifiers in activated sludge systems must be designed not only for hydraulic overflow rates
but also for solids loading rates. This is due mainly to the need for both clarification and
thickening in activated sludge clarifiers to provide both a well clarified effluent and a
concentrated return sludge.
When the MLSS concentration is less than about 3,000 mg/1, the clarifier size will normally
be governed by hydraulic overflow rates. At higher MLSS values, the ability of the clarifier
to thicken solids becomes the governing factor. Therefore, solids loading rates become more
6-3
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critical in determining tank size. Design size should be computed for both average and peak
conditions to insure satisfactory effluent quality at all times.
TABLE 6-2
TYPICAL DESIGN PARAMETERS FOR SECONDARY CLARIFIERS
Overflow Rate
Type of Treatment Average Peak
gpd/sq ft
Settling Following
Trickling Filtration 400-600 1,000-1,200
Settling Following Air-
Activated Sludge
(Excluding Extended
Aeration) 400-800 1,000-1,200
Solids Loading!
Average Peak
Ib solids/day/sq ft
Settling Following
Extended Aeration 200-400
800
20-30
20-30
< 50
< 50
Settling Following
Oxygen-Activated
Sludge with Primary
Settling 400-800 1,000-1,200
25-35
<50
10-12
12-15
12-15
12-15
Allowable solids loadings are generally governed by sludge settling characteristics associated
with cold weather operations.
Depth of clarifiers in activated sludge systems is extremely important. The depth must be
sufficient to permit the development of a sludge blanket, especially under conditions when
the sludge may be bulking. At the same time, the interface of the sludge blanket and the
clarified wastewater should be well below the effluent weirs.
For rectangular tanks, criteria similar to those given for primary tanks also apply to
secondary tanks. However, in long tanks it is common practice to locate the sludge
withdrawal hopper about 1/3 to 1/2 the distance to the end of the tank to reduce the
effects of density currents.
Typical design parameters for clarifiers in activated sludge systems treating typical domestic
wastewaters are also presented in Table 6-2. The design of these clarifiers should be based
upon an evaluation of average and peak overflow rates and solids loadings. That
combination of parameters which yields the largest surface area should be used.
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6.3 Physical Upgrading of Clarifiers
6.3.1 Optimization of Aeration Tank Clarifier Relationship
The activated sludge process is flexible in that the operator can vary the rate of sludge
wasting, thereby controlling the MLSS concentration to maintain the F/M and SRT at any
desired value. However, the MLSS in the aeration tank, in conjunction with the return
sludge rate, has a direct effect on the solids loading and capture efficiency of the secondary
clarifiers. Therefore when evaluating the performance of an existing treatment plant and/or
planned expansions, it is important to analyze aeration tanks, secondary clarifiers, and
return and waste sludge pumps as a system.
In many treatment plants, the MLSS concentration will be altered to achieve a desired
operating condition in the aeration tanks without considering the possible adverse effects on
the secondary clarifiers. It is common practice among many operators to carry high MLSS
concentrations to increase SRT, thereby oxidizing more organic matter and reducing waste
sludge mass. In many instances, the high solids loading rate associated with high MLSS
concentrations will cause the sludge blanket to rise to a level where solids will be swept over
the effluent weirs. Thus it is essential that a proper solids balance be maintained between
the aeration tank and clarifier.
Improved performance through optimization of internal process balances was accomplished
at both the Coldwater Creek Wastewater Treatment Plant in St. Louis, Missouri, and the
Sioux Falls, South Dakota, Wastewater Treatment Plant. Case histories of the upgrading of
these plants presenting the major modifications implemented and resultant removal
efficiencies are described in Chapter 13.
6.3.2 Inlet-Outlet Configurations
Typical inlet-outlet arrangements for circular clarifiers are shown on Figure 6-1. Figure 6-la
is the most common arrangement currently in use. The influent is carried to the center of
the tank by means of a pipe that may be either suspended a few feet below the water
surface or constructed beneath the tank floor. A circular baffle is provided at the center of
the tank to dissipate the energy generated by the influent discharging into the tank and to
distribute flow equally in all directions. The diameter of the influent well should be
approximately 15 to 20 percent of the clarifier diameter. The well should extend deep
enough into the clarification zone (at least 1/2 the tank depth) to insure good distribution,
but should be high enough above the sludge zone to prevent scouring of settled sludge. The
clarified effluent leaves the tank over a circular weir located at the perimeter of the tank.
For large circular tanks, the clarified effluent may be removed by means of a cantilevered
circular effluent trough at a point about 2/3 to 3/4 of the clarifier radius. The purpose of
such an arrangement is to reduce weir hydraulic loadings and approach velocities to
discourage the creation of currents which may travel along the clarifier floor and up the
6-5
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FIGURE 6-1
TYPICAL CLARIFIER CONFIGURATIONS
EFFLUENT
SLUDGE
INFLUENT
8-ta CIRCULAR CENTER-FEED CLARIFIER WITH
A SCRAPER SLUDGE REMOVAL SYSTEM
INFLUENT
EFFLUENT
7-* SLUDGE
6-lb CIRCULAR RIM-FEED, CENTER TAKE-OFF CLARIFIER WITH A
HYDRAULIC SUCTION SLUDGE REMOVAL SYSTEM
SLUDGE
B-lc CIRCULAR RIM-FEED, RIM TAKE-OFF CLARIFIER
6-6
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clarifier wall. However, many designs of this type have an opposite result when the
cantilevered trough is insufficiently far from the tank wall. In such cases, high upward
velocities in the annular space between the wall and the trough will actually create a
significantly worse condition than that which would have existed with a single peripheral
weir.
Rectangular tanks also perform effectively with appropriate inlet and outlet arrangements.
Frequently, poor hydraulic distribution in an influent channel common to several
rectangular tanks will cause some of the tanks to become overloaded. It is good design
practice to provide sufficient head loss in the ports feeding the tanks which will negate the
effect of water level variations within the distribution channel.
Inlet baffles for rectangular tanks should extend to at least 1/2 the tank depth to minimize
short-circuiting. Effluent weir arrangements should be checked carefully for excessive
upward velocities. A common tank deficiency is to have adequate weir length, but with
closely spaced troughs creating high rising velocities, enhancing solids losses at higher sludge
blanket levels.
Inefficient scum removal can adversely affect plant performance. Scum is usually trapped on
the tank surface by means of baffles placed in front of the effluent weirs. Skimmers
connected to the arms of the sludge collectors, separate helical devices, or tiltable slotted
pipes are used for scum collection and withdrawal. To insure adequate scum capture, baffles
should project 9 to 12 inches below the water surface. Since scum removal is ordinarily
accompanied by large volumes of water, decanting facilities are desirable to avoid excess
water in digestion and dewatering operations.
Two types of rim-feed circular clarifiers are shown on Figures 6-lb and 6-lc. A baffle
located a short distance from the tank wall forms an annular space into which the influent
wastewater is discharged in a tangential direction. The influent flows spirally around the
tank and underneath the baffle. The effluent may be removed by either a centrally located
weir trough as shown on Figure 6-lb or by a trough along the tank perimeter as shown on
Figure 6-lc. Typically, scum is removed by means of a slotted pipe at the end of the
influent annular ring.
It has been demonstrated that the flow pattern generated by the rim-feed clarifier is more
stable with varying hydraulic loads than the flow pattern which is developed in circular
center-feed clarifiers (1) (2) (3). Bergman (3) noted that rim-feed clarifiers with center
takeoff were more stable than center-feed clarifiers when subjected to varying thermal loads,
i.e., variation between the temperature of the incoming wastes and the wastes in the
clarifiers. Studies undertaken by Dague, et al, (2) indicate that center-takeoff rim-feed
clarifiers are hydraulically two to four times more efficient than center-feed clarifiers.
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Cleasby, et al, (4) indicate that rim-feed and center-feed clarifiers should have the same
removal efficiency when handling discrete particles and that rim-feed clarifiers should have
higher removal efficiencies when handling flocculent solids.
Although some of the papers cited above suggest that the rim-feed concept is superior to
center-feed units, the conversion from center-feed to rim-feed has not been used extensively
as an upgrading technique. Some of the impediments to such a conversion are:
1. Structural modifications to inlet and outlet configurations may be extensive,
especially if the inlet pipe passes under the tank floor.
2. Scum removal in rim-feed, center-takeoff clarifiers is somewhat less effective
because only the annular influent ring is available for scum separation and
removal. With this arrangement, floating sludge is free to escape the tank.
3. The location of the effluent weirs in rim-feed, center-takeoff clarifiers results in
relatively short weir length and high approach velocities.
4. The support system of the central weir trough must be designed to avoid
interference with existing sludge collector mechanisms.
The limitations outlined in the last three comments above are largely eliminated in rim-feed,
rim takeoff designs.
It is now more common to construct center-feed tanks as deep as originally used in rim-feed
units (12 to 14 feet). The trend is also towards deeper inlet baffles and to suction type
sludge collectors in center-feed tanks. These features have commonly been installed in
rim-feed tanks for many years. Consequently, the advantages of the rim-feed units have been
reduced. In most cases, it can be expected that the cost benefit of conversion from
center-feed to rim-feed will be lower than the cost benefit of other upgrading techniques.
Fall (5) has described a system in the Greater Peoria Sanitary District Sewage Treatment
Plant where changes in inlet and outlet designs were implemented to improve clarifier
performance. The existing primary clarifiers were square tanks with inlet ports along one
side and outlet ports along the opposite side. The secondary clarifiers (following aeration
tanks) were similar in design except an effluent weir was provided in lieu of outlet ports.
These tanks were experiencing extreme short-circuiting, up to about 1/4 of the theoretical
detention time, resulting in poor SS capture. One primary and one final clarifier were
modified as indicated on Figure 6-2. After modification, the influent enters the tank from
two opposite sides through pipes that discharge near the bottom of the tank. Velocities
generated by the influent pipes carried the solids to the center of the tank where the solids
are collected. The clarified effluent is removed over a peripheral weir. The modified tanks
showed actual and theoretical detention times that were about equal. Peak primary tank
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FIGURE 6-2
CLARIFIER MODIFICATIONS AT THE GREATER PEORIA
SANITARY DISTRICT SEWAGE PLANT
SLUDGE
EFFLUENT CONDUIT
EFFLUENT
TROUGH
EFFLUENT WEIR
PIPES
-INFLUENT
CHANNEL
INFLUENT
6-2a TOP VIEW
INFLUENT CHANNEL
.EFFLUENT TROUGH
SLUDGE
INFLUENT
PIPES
6-2D SIDE VIEW
6-9
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overflow rates as high as 4,100 gpd/sq ft were encountered with very little loss in average SS
removal efficiency. Based on these test results, the entire plant was converted to the
modified clarifier design. Since this conversion, there have been other process modifications
and changes in plant loadings. As a result, it has been difficult to verify the improvement in
plant performance due to the clarifier modifications.
6.3.3 Control of Problem Sludges
Two of the major problems encountered in activated sludge systems are rising sludge and
bulking sludge. In many instances these problems can be corrected through operational
control. Where operational control is limited, plant modifications may be required.
6.3.3.1 Rising Sludges
Sludge accumulation on the bottom of secondary clarifiers for excessively long periods
often causes oxygen deficient conditions. If the biological system is nitrifying, the
accumulated sludge may undergo denitrification, which occurs when the oxygen contained
in nitrites and nitrates is used to satisfy the oxygen demand of the settled sludge. As
nitrate-bound oxygen is consumed, nitrogen gas is formed and becomes trapped in the
sludge mass. If enough gas is formed, portions of the sludge mass become buoyant and rise
to the surface.
The problem can usually be solved by reducing the residence time of the sludge in the
clarifier. Corrective actions that may be taken are:
1. Increase the return sludge pumping rate.
2. Increase the speed of the sludge collector (if possible).
\
3. Increase the sludge wasting rate.
In some instances, floating sludge has been caused by the organism Nocardia actinomyces.
This condition is difficult to eliminate and is evidenced by a heavy brown scum on aeration
tanks and final clarifiers. Some success has been achieved by the following:
1. Chlorination of return sludge or MLSS
2. Lowering DO to 0.5 mg/1 or less
3. Reducing MLSS concentration
4. Raising pH to 7.5 to 8.0 by lime addition (6).
6-10
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Plant modifications that may improve sludge management include increasing the capacity of
return and waste sludge pumping, installing sludge blanket finders, and replacing plow-type
sludge collectors with suction-type units. The suction-type units will remove the sludge
directly from the clarifier bottom through suction orifices that sweep the entire bottom of
the tank during each revolution. They are recommended for clarifiers with diameters greater
than 50 feet.
6.3.3.2 Bulking Sludges
A bulked sludge is one which has poor settling characteristics and poor compactibility. Two
principal types of sludge bulking have been identified. One is caused by the growth of
filamentous organisms (7) (8) or organisms which can grow in a filamentous form under
adverse conditions. The other is caused by bound water, in which the bacterial cells
composing the floe swell through the addition of water to the point that their density
approaches that of water.
The causative factors normally associated with sludge bulking and the recommended
preventive actions are as follows:
1. DO content: Limited DO has been noted more than any other cause of bulking.
The DO in the aeration tank should be normally maintained at or above 2 mg/1.
2. Process loading: The F/M should be maintained within the recommended range
for the operational mode of the type of activated sludge plant being used.
Recommended ranges are given in Chapter 5. SRT has also been shown to be an
effective tool in preventing the growth of filamentous organisms. Stamberg et al
(9) found, for oxygen and conventional activated sludge systems at the EPA-DC
Pilot Plant, that for SRT's below five days, both systems exhibited filamentous
growth. Such growths did not occur at SRT's greater than five days. When
encountering filamentous growth for a few days only, increasing the SRT
reestablished a filamentous free sludge in several days.
3. Wastewater characteristics: If it is known that industrial wastewaters are being
introduced to the system, the quantity of nitrogen and phosphorus in the
wastewater should be checked, since limitations of either are known to favor
bulking. Wide fluctuations in BOD and pH should also be checked. The presence
of toxic substances has also been implicated as a cause of bulking.
4. Clarifier operation: Evaluation of clarifier performance should be made. This is
particularly applicable to rectangular and center-feed circular tanks. Profile
sampling of the sludge blanket may show that a large part of the sludge is retained
in the tank for many hours rather than the desired 30 minutes. Such a condition
may require major modifications to the clarifier's sludge collection system.
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When filamentous growths become firmly entrenched in the biomass, it is doubtful any of
the above actions will be effective. In these situations, chemical treatment to selectively
destroy these growths has been successful as discussed below:
1. Chlorination of wastewater, mixed liquor and return sludge has been found
effective in controlling bulking due to filamentous growths. It is ineffective when
bulking is due to light, diffuse floes. Recommended dose rates based on return
sludge flow are between 0.2 and 1.0 Ib of chlorine per 100 Ib of return sludge
solids. This method is only considered as a temporary measure and should not
exceed 24 hours duration. Selective destruction of filamentous organisms by
chlorination generally will cause a turbid effluent until the destroyed organisms
are washed out of the system.
2. Application of hydrogen peroxide has also been employed successfully for control
of filamentous growths for both air and oxygen activated sludges. At the EPA-DC
Pilot Plant, continuous dosing of the return sludge for 24 hours with a hydrogen
peroxide dosage of 200 mgA (based on plant influent) resulted in an immediate
improvement in sludge settling and a gradual recovery of system
performance (10).
6.3.4 Tube Settlers
The information regarding tube settlers presented in this section is addressed only to the
upgrading of existing, conventionally designed ciarifiers. A detailed discussion for the design
of tube settlers is presented in the Process Design Manual for Suspended Solids
Removal (11).
According to the classical theory of discrete particle settling, the efficiency of suspended
particle removal in a sedimentation basin is solely a function of overflow rate and is
independent of depth and detention time. If the above theory is applicable to raw
wastewater or activated sludge floe settling, then the clarifier performance could be
improved by introducing a number of trays or tubes in the existing ciarifiers. This concept
has led to the development of the tube settler. Although tube settlers have been installed in
several existing plants, little comparative data have been published to verify their
effectiveness as an upgrading technique. However, it can be expected that as additional
operational data are developed, the use of tube settlers may warrant further consideration.
Tube settlers have been used in primary and secondary ciarifiers to improve performance as
well as to increase throughput in existing ciarifiers. Conley and Slechta (12) and Hansen, et
al, (13) have described the performance of several plant-scale installations of tube settlers in
primary and secondary ciarifiers. The results of their studies indicate that the overflow rates
in primary ciarifiers can be substantially increased while producing the same quality effluent
as the control unit without the settlers.
6-12
-------
One manufacturer of tube settlers has suggested the peak overflow rate for circular
secondary clarifiers in activated sludge systems be limited to about 1.0 gpm/sq ft
(1,440 gpd/sq ft) for the tube settler area (14). These values apply to 70 deg F wastewaters
and should be reduced by a factor of two at 40 deg F. This value is essentially the same as
that recommended for secondary clarification without tube settlers. Therefore, tube settlers
are not expected to accomplish a saving in tank area. A minimum tank depth of 10 to
12 feet is necessary to accommodate the tube modules and to provide sufficiently low
approach velocities.
Tube settlers enhance the ability to capture settleable solids at high overflow rates because
the depth of settling has been reduced to a few inches in the tube. It should be realized that
tube settlers do not improve the efficiency of primary clarifiers that are already achieving
very high (40 to 60 percent) removals of SS. Moreover, tube settlers will neither remove
colloidal solids that remain in suspension nor induce additional coagulation to effect added
particle removal. A portable tube settler unit has been evaluated following primary clarifiers
that were handling paper mill wastewaters which contain a high concentration of solids.
With this arrangement, it was possible to appreciably increase the hydraulic capacity of the
primary system without reducing the quality of the effluent (15).
The characteristics of the existing tanks will play a large part in determining the feasibib'ty
of installing tube settlers. In many upgrading situations, the tube settlers are installed over a
portion of the basin near the outlet. For example, in circular basins, the tubes are placed in
pie-shaped segments in a ring around the basin outer wall. Major physical factors that must
be considered are:
1. Provision of adequate tank depth
2. Provision for adequate sludge management and storage
3. Modifications to existing effluent weir troughs
4. Potential conflicts with sludge and scum collection equipment.
Fouling due to biological growths on the tubes has been cited as an operating problem. This
problem has largely been overcome by the installation of permanent air headers beneath the
entrance to the tubes. Typical installation details are shown in the Process Design Manual
for Solids Removal (11). Algal growths near the water surface can sometimes be controlled
by covering the clarifiers to block direct sunlight.
Since the flocculating and settling characteristics of sludge vary from plant to plant, each
case should be evaluated separately for suitable design criteria. Small pilot units are available
from the manufacturer for this purpose.
6-13
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Operational data for tube settlers at three locations are presented in Table 6-3. These
installations employ tube settlers for clarifying activated sludge mixed liquor. At the two
locations, Hopewell Township and Lebanon, where comparative data are available, the tube
settlers have markedly outperformed conventional clarification. Case histories for these two
installations are presented in the Process Design Manual for Solids Removal (11).
6.4 Chemical Treatment
Chemical addition to primary and secondary clarifiers in this manual is addressed only to
increased solids and BOD removal. Chemical treatment for phosphorus removal is covered in
detail in the Process Design Manual for Phosphorus Removal (18). A detailed discussion of
chemical properties, storage facilities, and feed equipment is presented in the Process Design
Manual for Suspended Solids Removal (11).
6.4.1 Chemicals Used
The chemicals commonly used in wastewater treatment are the salts of iron and aluminum,
lime, and synthetic organic poly electrolytes. The iron (ferrous and ferric) and aluminum
salts (sodium aluminate or alum) react with the alkalinity and soluble orthophosphate in
wastewater to form precipitates of the respective metallic hydroxides or phosphates. In
addition, they destabilize the colloidal particles that would otherwise remain in suspension.
These precipitates, along with the destabilized colloids, flocculate and settle readily in a
clarifier.
Both alum and sodium aluminate exhibit great capability for total phosphorus removal, but
the use of alum introduces six times as much dissolved solids to the wastewater as does
sodium aluminate (19). Normally, lime is used to precipitate hydrous oxides of iron and
aluminum when the alkalinity of wastewaters is low. The reaction of iron and aluminum
salts is pH-dependent and has to be evaluated for each case to determine the most effective
pH range and the optimum chemical dosage.
The addition of hme alone is also effective in coagulating wastewater. The positive calcium
ions destabilize colloidal particles while precipitating soluble orthophosphates as
hydroxyapatite. Since lime treatment takes place at pH 9.0 to 11.5, pH adjustment may be
required before subsequent biological treatment depending on the operating pH and the
degree of biologically induced recarbonation that occurs. This requirement may be reduced
or eliminated for biological systems that nitrify.
6.4.2 Addition of Chemicals to Primary Clarifiers
In the early stages of wastewater treatment, especially in the late 1930's and early 1940's,
chemicals were used to improve the efficiency of primary clarification systems. Later, when
these systems were followed by secondary treatment processes, the practice of adding
6-14
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TABLE 6-3
TUBE SETTLER INSTALLATIONS
Operational Data Using Tube Settlers
Plant Location
Hopewell Township ,
Pennsylvania
£ Trenton,
01 Michigan
Lebanon,
Ohio
Type
Activated
Sludge
Activated
Sludge
Activated
Sludge
Plant Flow Tube Existing Facility Tube Over- Tank Over-
Design Actual Location Overflow Rate Effluent SS flow Rate flow Rate Effluent SS
mgd mgd gpm/sq ft mg/1 gpm/sq ft gpm/sq ft mg/1
0.13 0.13 Secondary 0.34 60-70 -
Clarifier - -2 0.68 27
6.5 5.6 Secondary - - 0.56 0.29 8
Clarifier
0.75 1.25 Secondary 0.61 61 0.85 0.61 30
Clarifier
Reference
13
16
17
-------
chemicals to upgrade primary treatment became unnecessary. However, the technique of
adding chemicals to the primary clarifier is still an effective upgrading procedure for a
secondary plant when the following conditions exist (20):
1. Wastewater flow is intermittent or varies greatly.
2. Space available for additional clarification facilities is limited.
3. Industrial wastes that would interfere with biological treatment are present.
4. Plant is hydraulically and/or organically overloaded.
5. Improvements in existing treatment performance are required as an interim
measure before the addition of new facilities.
When considering the addition of chemicals to primary clarifiers, it is important to examine
the effect of increased primary clarifier efficiency on subsequent treatment units. The
increased removal of SS and BOD from raw waste water can affect the downstream
biological process in several ways. If the BOD load to the aerator falls below 0.25 to
0.35 Ib BOD/lb MLVSS/day for extended periods of time, nitrification conditions can
develop in the aerator. This can reduce the total oxygen demand of the effluent, but will
impose an added oxygen demand on the aeration facility because the oxidation of one
pound of ammonia nitrogen requires about 4.6 pounds of oxygen.
A decrease in loading to the aerator will normally require more careful management of
sludge to ensure stable operation of the aeration basin. However, the quantity of excess
activated sludge generated under these reduced loading conditions will be substantially less
than that generated under normal loading conditions. This may be considered an added
advantage of adding chemicals to the primary clarifier, if the additional fines captured
chemically do not seriously degrade the quality of the primary sludge.
Little information has been generated regarding the periodic addition of chemicals to the
primary clarifier for controlling peak organic or hydraulic loads. This approach, while not
always applicable, can frequently be used to' maintain system stability during temporary
overload.
The effect of polyelectrolyte and coagulant addition on primary clarifier performance is
shown in Table 6-4. Where plant information is available, the performance of clarifiers
before and after the addition of chemicals is shown. The data show a considerable variation
in BOD and SS removals. However, the effect of polyelectrolyte addition is pronounced
where the existing clarifier performance is poor. The variations shown make it difficult to
project the expected improvement due to chemical addition at a specific treatment plant.
6-16
-------
TABLE 6-4
EFFECT OF CHEMICAL TREATMENT ON PRIMARY CLARIFIER PERFORMANCE
Type and Amount
of
Chemical Added
Purifloc - A21 (0.95 mg/1)
DOW -SA 1193 (0.2 mg/1)
Purifloc - A21 (1 mg/1)
Purifloc - A21 (0.75 mg/1)
Purifloc - A21 (0.89 mg/1)
DOW - SA1193 (0.25 mg/1)
Purifloc - A21 (1 mg/1)
FeCl2 + NaOH + Purifloc - A23 (0.3 mg/1)
FeCl2 + NaOH + Purifloc - A23 (0.3 mg/1)
Purifloc - A21 (1 mg/1)
Purifloc - A23 (0.25 mg/1)
FeCl3 + Purifloc - A23
FeCls + Purifloc - A23
Purifloc - A21 (0.74 mg/1)
Purifloc - A21M (1.14 mg/1)
FeCls (20 mg/1) + Purifloc - A23 (0.3 mg/1)
FeCl3(35 mg/1 Fe3+) + Purifloc - A23 (0.5 mg/1)
FeCl3 (15-18 mg/1 Fe3+) + Purifloc - A23
(0.5 mg/1)
mg/1 Fe3+) + Purifloc - A23 (0.5 mg/1)
0-25 mg/1 Fe3+) + Purifloc - A23
(0.4 mg/1)
FeCl3(22 mg/1 Fe3+) + Purifloc - A23 (0.5 mg/1)
Alum (15-20 mg/1 A13+) + Purifloc - A23
(0.5 mg/1)
Alum (90 mg/1) + Polyelectrolyte (0.4 mg/1)
Alum (110 mg/1) + Polyelectrolyte (0.35 mg/1)
Weight
Performance Preceding Chemical Treatment Ratio of
SS Removed
mg/1
13
13
157
26
113
120
107
230
104
52
93
93
1
-
-
-
percent
12
12
43
18
43
47
47
82
49.7
31
33
33
50
43
1.3
-
35.5
-
-
_
BOD
mg/1
28
28
82
_
50
_
135
111
83
_
47
53
53
_
_
_
-
-
-
_
Removed WAS/PS1
percent
26 0.61
26 0.61
23
_
22
0.8
37
31
43.8
0.79
31 1.44
34
34
36
_ _
_ _
-
19.1
-
-
_ _
Performance After Chemical Treatment
SS Removed
mg/1
75
72
281
69
159
151
169
379
173
80
196
213
38
323
_
177
41
61.7
134.8
157
204
percent
65
55
76
52
60
61
62
79
76.8
51
74
68
63
63
24.4
80
63.6
74.5
74.0
84
70
84
74
BOD Removed
mg/1
46
36
127
-
87
-
154
74
105
-
58
102
97
-
249
-
115
226
423.9
66
126
percent
48
37
33
37
-
46
39
57.8
-
46.4
61
53
45
-
-
61
54.4
-
57.4
38
32
61.1
71
Weight
Ratio of
WAS/PS1 Reference
0.31 21
0.41 21
21
21
21
0.46 21
21
22,23
22,23
0.28 24
0.67 25
26
26
27
27
28
29
30
31
32
33
33
34
35
IWAS - Waste activated sludge
PS - Primary sludge
-------
For proper evaluation, the characteristics of the wastewater and the effects of recycle flows
from the sludge processing operations must be known. Such information can be obtained
through pilot plant tests or full-scale plant trials.
Freese, et al, (27) studied the application of polyelectrolytes for raw wastewater
flocculation in the District of Columbia 7/ater Pollution Control Plant. Full-scale plant trials
were conducted both with and without recycle of elutriate to the primary clarifiers. The
overflow from gravity thickening was recycled to the primary tanks for all tests. Without
elutriate recycle, primary tank BOD removals were increased from 36 to 45 percent and
SS removals from 50 to 63 percent. Overall BOD removal was improved from 74 to
78 percent, but no improvement in overall SS removal was obtained. When the elutriate was
added to the primary tank influent without poly electrolyte addition, poor capture of the
elutriate solids occurred. Although capture of elutriate solids improved when
polyelectrolyte was added, a major portion of these solids still escaped and were captured in
secondary treatment. This led to a gradual accumulation of elutriate solids in the solids
processing system, which increased the loadings on thickening, digestion and elutriation
tanks. The increased loadings on the elutriation tanks caused by recycled solids caused a
gradual decrease in capture efficiency, increasing the solids recycled to primary treatment. It
was concluded that polyelectrolyte application to the raw wastewater with elutriate recycle
would not prevent the buildup of large quantities of recycled solids through the solids
processing components, and that the elutriation tank efficiency would be the focus of
further tests.
It has since been found that polyelectrolyte addition to the elutriation tank has increased
the capture of solids in the elutriation process from 57 to 92 percent. The resulting
improvement in elutriate clarity has enabled the plant to tolerate the recycle without the
accumulation of fine solids (36).
Table 6-5 summarizes data of the effect of polyelectrolyte addition in primary clarifiers on
the primary and overall plant efficiency. The data reported covers both activated sludge and
trickling filter treatment plants.
Schmidt and McKinney (39) studied phosphorus removal by lime addition to the primary
clarifier of an activated sludge treatment plant. In this study, the system was operated at a
pH value of 9.5 which, during biological treatment, was reduced to a value between 7 and 8.
Therefore, no neutralization was required. The lime precipitation step reduced the BOD by
60 percent, SS by 90 percent, and total phosphorus by 80 percent. However, Schmidt and
McKinney indicated that the lime-primary sludge was gelatinous in nature and required
polyelectrolyte treatment prior to dewatering by vacuum filtration. They further indicated
that the mass of primary sludge is about twice that obtained by conventional settling,
although the total mass of primary and secondary sludge produced is increased by less than
50 percent. Lime addition to primary clarifiers for phosphorus removal has been used in
many locations. In all cases, significant improvements in both SS and BOD removal were
noted. Table 6-6 presents the results of some of these studies.
6-18
-------
TABLE 6-5
POLYELECTROLYTE ADDITION TO PRIMARY CLARIFIERS
Primary Clarifier
Total Plant
Percent Percent Percent Percent
Removal Removal Removal Removal
Before After Before After
Poly- Poly- Poly- Poly-
electrolyte electrolyte electrolyte electrolyte
Treatment Process Coagulant Dose Addition Addition Addition Addition Reference
Activated Sludge Purifloc
A-21
mg/1 BOD SS BOD SS BOD SS BOD SS
1 26-48-83-90-37
Trickling Filter Purifloc
A-21 1 23 43 33 76 79 72 85 84 37
Activated Sludge Purifloc
A-23 0.21 31 31 46 51 79 85 83 89 38
*Case history presented in Chapter 13.
TABLE 6-6
LIME ADDITION TO PRIMARY CLARIFIERS
Percent Removal Percent Removal
in Primary Before in Primary After
Location Lime Added Lime Addition Lime Addition Remarks Reference
mg/1 CaO BOD
Duluth,
Minnesota
Rochester,
New York
Lebanon,
Ohio
Richmond
Hill, Ontario
Central Contra
Costa, Calif.
75 50
125 55
100
145
175 21
378 46
303 37
SS BOD
70 60
70 75
50
66
37 71
71 74
71 69
SS
75
90
80-90 Jar tests
74 Pilot plant
77 Full-scale plant
79 Full-scale test
76 Full-scale test
18
18
40
41
42
6-19
-------
As mentioned above, the addition of lime to the primary clarifier can be expected to
increase the primary sludge mass to about twice that obtained by conventional primary
settling, depending on the operating pH and alkalinity of the incoming wastewater.
Therefore, a complete evaluation of the sludge handling facilities must be made when
considering this technique. For instance, some states have cautioned against this practice
when the primary sludge is to be anaerobically digested.
6.4.3 Use of Chemicals in Secondary Processes
Much of the published information available on the addition of chemicals to secondary
processes emphasizes their use for phosphorus removal. These studies have shown that in
many cases the addition of iron and aluminum salts can significantly improve secondary
clarifier performance depending upon the applied dosage, the point of addition, and the
flocculent nature of the biomass. The results of several of these studies are presented in
Table 6-7.
In certain processes, such as trickling filtration and extended aeration, solids may not
flocculate and settle well in the secondary clarifier. In these instances, the addition of iron
or aluminum salts will provide a greater benefit in improving plant performance than would
a similar chemical addition to an activated sludge that flocculates and settles well.
Polyelectrolytes have also been used to improve the performance of secondary clarifiers.
Singer, et al, (43) studied the effect of adding cationic and anionic polyelctrolytes to
improve settling characteristics of bulking activated sludge in the laboratory. Their studies
indicated that cationic polyelectrolytes at a concentration of 2.0 to 3.0 mg/1 were effective
in coagulating a bulking activated sludge but that the anionic polyelectrolyte tested had no
beneficial effect on improving settling. Goodman and Mikkelson (53) on the basis of
full-scale studies, concluded that application of cationic polyelectrolytes to primary clarifier
effluent at the rate of 0.1 Ib/ton of secondary dry solids increased overall BOD removal
efficiency to 95 percent and decreased the loss of solids in the secondary effluent of the
activated sludge plant.
Based on studies conducted at the Hanover Treatment Plant by the Metropolitan Sanitary
District of Chicago, Zenz and Pivnicka (54) have shown that the addition of alum to
aeration tanks (primarily intended for soluble phosphorus removal) improved flocculation
of activated sludge. However, their results indicated that increasing amounts of alum floe
escaped through the final clarifiers as the dosage of alum increased from an A1:P weight
ratio of 1.54 to 1.85. The addition of alum to the aeration tank favored the development of
lower organisms, while the higher forms such as protozoa and metazoa were absent.
Although higher forms of organisms are adversely affected by the addition of alum, BOD
removal is not affected (54) (55). Earth and Ettinger (47) found that dosages of 10 mg/1 of
alum (as A13+) in the secondary aerator did not interfere with the nitrification process.
Zenz and Pivnicka (54) also indicated that alum sludges can be stabilized anaerobically, and
that in this process the precipitated phosphate is not released, and is therefore permanently
removed from the system.
6-20
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TABLE 6-7
EFFECT OF CHEMICAL TREATMENT ON SECONDARY CLARIFIER PERFORMANCE
Effluent BOD5 Effluent SS Effluent BOD5 Effluent SS Total
Location
Richardson, Texas
Chapel Hill,
North Carolina
Pennsylvania State
Cincinnati,
Ohio
Lebanon,
Ohio
Minneapolis,
Minnesota
Madison,
Wisconsin
Lmversity Park,
Pennsylvania
Bloomington,
Illinois
Blue Plams,2
Washington, D.C
Sandusky, Ohio
Michigan City,
Indiana
Guelph, Ontano
Palmetto, Florida
Type of Plant
Tnckling filter
std. rate
Tnckling filter
high rate
Conventional
activated sludge
Activated sludge
(100 gpd pilot)
Activated sludge
(0.11 mgd pilot)
Tnckling filter
low rate
Trickling filter
Activated sludge
Tnckling filter
Modified Activated
sludge
Conventional
activated sludge
Conventional
activated sludge
Conventional
activated sludge
Tnckling filter
Location of Chemical and (or COD) Before Before (or COD) After After Phos
Chemical Addition Dosage Chemical Addition Chemical Addition Chemical Addition Chemical Addition Removal Reference
Before final settling
Before final settling
Aerator effluent
Aerator
Final clanfier
Before final settling
Before final settling
Before final settling
Aerator
Before final settling
Before final settling
Aerator
Aerator
Aerator
Before final settling
mg/1 mg/1
Al/P Mole
Dosage 1.6/1 20 15
Al/P Mole
Dosage 1.6/1 44 64
AVPwt
Ratio 3/1 13 26
10mg/IAl3+ (89%) 1 (95«)1
Add lime to
raisepH=9.4-10.9 435
720 mg/1
Ca(OH)2 83
200 mg/1
Alum 8-29
160 mg/1
Alum 18 31
46 mg/1
Na2Al2O4 61 95
33 9 mg/1 Fe3+
+0.7 mg/1
Punfloc - A23 8 8 12 7
25-30 mg/1 Fe3+
+0.5 mg/1
Punfloc -A23 130 496
26 mg/1 Alum 47 48
50 mg/1 Alum 38 39
60 mg/1 Alum 68 53
80 mg/1 Alum 46 41
89 mg/1 Alum 50 57
50 mg/1
Alum 9 24
60 mg/1
Alum 13 19
100 mg/1
Alum 26 38
45 mg/1 30-40
Alum
mg/1 mg/1 percent
<5 <7 95 44
15 34 82 45
9 22 86 46
(92%)' (96%)! 94 47
16.5 48
27 86 49
1 8-2.9 98 7 50
6 19 96 4
51
23 8 93 4
50 8.6 50
33 16.0 50
40 43
27 36
25 36 52
41 31
30 31
2 15 80 35
9 7 922 35
14 22 87 35
10 35
1 Percent removal
^Data arc monthly average.
6-21
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Derrington, et al, (44) have reported adding 435 gpd of liquid alum (8.3 percent A^Os) to
the secondary clarifier of the Richardson, Texas, trickling filter plant treating 1.6 mgd of
wastewater. The results indicated a reduction in effluent BOD and SS concentrations from
20 mg/1 and 15 mg/1, respectively, to 5 mg/1 and 7 mg/1. The effluent phosphorus
concentration was reduced from 8 mg/1 to 0.5 mg/1. In addition, Derrington, et al, (44) have
reported problems of reduced alkalinity in sludge undergoing digestion when alum was used
as coagulant in primary treatment at the Richardson, Texas, plant. The addition of alum to
raw wastewater was discontinued after nine days total operation to prevent further
reduction in alkalinity and pH in the digester. Studies undertaken at Pomona, California
(46) indicated that the turbidity of the final effluent may be increased when alum is added
to the aeration system. This may occur when the alum dosage is sufficient to lower the pH
below the optimu m pH for good alum flocculation, resulting in an increase in fines in the
final effluent.
Lime addition may not be feasible for upgrading activated sludge secondary clarifiers
because of the potential adverse effect of recirculated lime sludge on mixed liquor microbial
characteristics. Lime addition to either trickling filter or activated sludge secondary clarifiers
will require pH adjustment of the effluent before discharge to the receiving waters. Lime
addition to primary clarifiers may be used, if consideration is given to controlling the pH
within acceptable limits for the subsequent processes, and to changes in sludge
characteristics and handling requirements. Generally, chemical addition will increase the
weight of solids and/or the volume of sludge. Accordingly, sludge piping, pumping and
process units should be of sufficient size and capacity to handle the increased quantities of
sludge.
6.5 References
1. Katz, W. J., and Geinopolos, A., A Comparative Study of the Hydraulic Characteristics
of Two Types of Circular Solids Separation Basins. In Biological Treatment of Sewage
and Industrial Wastes, Vol. II, Anaerobic Digestion and Solids-Liquid Separation. Paper
presented at the Conference on Anaerobic Digestion and Solids Handling, April 1957,
Manhattan College. Edited by B. J. McCabe and W.W. Eckenfelder, Jr., New
York: Reinhold Publishing Corporation, pp. 196-206 (1958).
2. Dague, R.R., and Bauman E.R., Hydraulics of Circular Settling Tanks Determined by
Models. Paper presented at the 1961 Annual Meeting of Iowa Water Pollution Control
Association, Lake Okoboji, Iowa (June 8, 1961).
3. Bergman, B.S., An Improved Circular Sedimentation Tank Design. Journal Institute
Sewage Purification, Part I, pp. 50-67 (1958).
4. Cleasby, J.L., Bauman, E.R., and Schmid, L., Comparison of Peripheral Feed and
Center Feed Settling Tanks Using Model. Progress Report to Lakeside Engineering
6-22
-------
Corporation, Engineering Experiment Station Project 387-S, Iowa State University,
Ames, Iowa (February, 1962).
5. Fall, E.B., Jr., Redesigning Existing Treatment to Increase Hydraulic and Organic
Loading. Presented at the 43rd Annual Conference of the Water Pollution Control
Federation, Boston, Massachusetts (October, 1970).
6. Hankin, L., Glover, W.D., and Anagnostakis, S.L., Elimination of Heavy Biological
Scum in Secondary Sedimentation Tanks. Journal of the New England Water Pollution
Control Association, Vol. 6, No. 1, pp. 80-84 (June, 1972).
7. Farquhar, G.J., and Boyle, W.C., Identification of Filamentous Microorganisms in
Activated Sludge. Journal Water Pollution Control Federation, 43, No. 4, pp. 604-622
(April, 1971).
8. Farquhar, G.J., and Boyle, W.C., Occurrence of Filamentous Microorganisms in
Activated Sludge. Journal Water Pollution Control Federation, 43, No. 5, pp. 779-798
(May, 1971).
9. Stamberg, J.B., Bishop, D.F., Hais, A.B., and Bennett, S.M., System Alternatives in
Oxygen Activated Sludge. Presented at the 45th Annual Conference of the Water
Pollution Control Federation, Atlanta, Georgia (October, 1972).
10. Cole, C.D., Stamberg, J.B., and Bishop, D.F., Hydrogen Peroxide Cures Filamentous
Growth in Activated Sludge. Journal Water Pollution Control Federation, 45, No. 5,
pp. 829-836 (1973).
11. Process Design Manual for Suspended Solids Removal.U. S. Environmental Protection
Agency, Office of Technology Transfer, Washington, D.C. (1974).
12. Conley, W.R., and Slechta, A.F., Recent Experiences in Plant Scale Application of the
Settling Tube Concept. Presented at the 43rd Annual Conference of the Water
Pollution Control Federation, Boston, Massachusetts (October, 1970).
13. Hansen, S.P., Gulp, G.L. and Stukenberg, J.R., Practical Application of Idealized
Sedimentation Theory in Wastewater Treatment, Journal Water Pollution Control
Federation, 41, No. 8, pp. 1421-1444 (1969).
14. Neptune Microfloc Incorporated, Application Criteria for Tube Settling in Activated
Sludge Plant, Secondary Chrifiers. Technical Release No. 3 (1972).
6-23
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15. Abrams, R.C., Use of Inclined Tube Settlers for Improved Clarifier Performance. Paper
presented at the 32nd Annual Shibley Award Meeting of the Pacific Section, TAPPI,
Tacoma, Washington (March 31, 1970).
16. Neptune Microfloc Incorporated, City of Trenton Sewage Treatment Plant. Case
History No. 27 (1971).
17. Oppelt, E.T., Evaluation of High Rate Settling of Activated Sludge. Interim U. S. EPA
Internal Report, Advanced Waste Treatment Laboratory, Cincinnati, Ohio (1973).
18. Process Design Manual for Phosphorus Removal. U. S. Environmental Protection
Agency, Office of Technology Transfer, Washington, D.C. (1974).
19. .Brenner, R.C., Phosphorus Removal by Mineral Addition. Nutrient Removal and
Advanced Water Treatment Symposium. Sponsored by Federal Water Pollution
Administration, Cincinnati, Ohio (April 29-30,1969).
20. Sewage Treatment Plant Design. Water Pollution Control Federation Manual of
Practice No. 8, Washington, D.C. (1959).
21. Anon, Effects of Raw Sewage Flocculation in Secondary Waste Treatment Plants.
Midland, Michigan: The Dow Chemical Co.
22. Wukasch, R.F., The Dow Process for Phosphorus Removal. Paper presented at the
Phosphorus Removal Symposium. Sponsored by Federal Water Pollution Control
Administration, Chicago, Illinois (June, 1968).
23. Wukasch, R.F., New Phosphate Removal Process. Water and Wastes Engineering, 5,
No. 9, pp. 58-60(1968).
24. Voshel, D., and Sak, J.G., Effect of Primary Effluent Suspended Solids and BOD on
Activated Sludge Production. Journal Water Pollution Control Federation, 40, No. 5,
Part 2, pp. R203-R212 (1968).
25. Wirts, J.J., The Use of Organic Polyelectrolyte for Operational Improvement of Waste
Treatment Processes. Federal Water Pollution Control Administration, Grant
No. WPRD 102-01-68 (May, 1969).
6-24
-------
26. Applications of Chemical Precipitation Phosphorus Removal at the Cleveland Westerly
Wasteimter Treatment Plant. Prepared for the City of Cleveland, Ohio, by the Dow
Chemical Co., Midland, Michigan (April, 1970).
27. Freese, P.V., Hicks, E., Bishop, D.F., and Griggs, S.H., Raw Wastewater Flocculations
with Polymers at the District of Columbia Water Pollution Control Plant. Federal
Water Quality Administration, Contract No. WPRD 53-01-67.
28. Schuessler, R.G., Phosphorus RemovalA Controllable Process. Paper presented at the
41st Meeting, Water Pollution Control Association of Pennsylvania, University Park,
Pennsylvania (August, 1969).
29. Parker, D.G., Raw Sewage Flocculation Trial at Stevens Point, Wisconsin, December 8,
1971 to March 4, 1972. The Dow Chemical Company Report (1972).
30. The Dow Chemical Company, Phosphorus Removal by Chemical Precipitation at the
Buffalo Sewage Treatment Plant. Report prepared for Consoer, Townsend and
Associates, Chicago and the Buffalo Sewer Authority (December, 1970).
31. Parker, D.G., Phosphorus Removal Trial, Lansing, Michigan. Report prepared for
McNamee, Porter, and Seeley Ann Arbor and the City of Lansing by the Dow
Chemical Co. (December, 1970).
32. Parker, D.G., Phosphorus Removal Trial, East Lansing, Michigan. Report prepared for
Hubbell, Roth, and Clark, Inc., and the City of East Lansing by the Dow Chemical Co.
(January, 1971).
33. Parker, D.G., Phosphorus Removal Trial, Fond du Lac, Wisconsin. Report prepared for
John A. Strand & Associates and the City of Fond du Lac, Wisconsin by the Dow
Chemical Co. (July, 1971).
34. Allied Chemical Canada, Ltd., Phosphorus Removal in a Primary Plant at Windsor,
Ontario. Wastewater News, Allied Chemical Canada, Ltd. (February 5,1973).
35. Ockershausen, R.W., Upgrading Wastewater Plant Effluent by Chemical Treatment.
Presented at the Joint Conference AWWA-FPCA, Orlando, Florida (November, 1973).
36. Dahl, Zelinski, and Taylor, Polymer Aids in Dewatering and Elutriation. Journal Water
Pollution Control Federation, 44, No. 2, pp. 201-211 (1972).
37. Mogelnicki, S., Experiences in Polymer Applications to Several Solids - Liquids
Separation Process. Proceedings - Tenth Sanitary Engineering Conference - Waste
Disposal from Water and Wastewater Treatment Processes, University of Illinois
(February 6-7, 1968).
6-25
-------
38. Wirts, John J., The Use of Organic Polyelectrolyte for Operational Improvement of
Waste Treatment Processes. Prepared for Federal Water Pollution Control
Administration, Grant No. WPRD 102-01-68 (May, 1969).
39. Schmidt, L.A., and McKinney, R.E., Phosphate Removal by a Lime-Biological
Treatment Scheme. Journal Water Pollution Control Federation, 41, No. 7,
pp. 1,259-1,279 (1969).
40. Villiers, Ronald V., Municipal Wastewater Treatment by Single Stage Lime
Clarification and Activated Carbon. Internal U. S. EPA paper, Advanced Waste
Treatment Research Laboratory, Cincinnati, Ohio (1971).
41. Black, S.A., and Lewandowski, W., Phosphorus Removal by Lime Addition to a
Conventional Activated Sludge Plant. Division of Research Publication No. 36, Ontario
Water Resources Commission (November, 1969).
42. Central Contra Costa Sanitary District - Project Report for the Water Reclamation
Plant. Brown and Caldwell Consulting Engineers, San Francisco, California (November,
1971).
43. Singer, P.C., Pipes, W.O., and Hermann, E.R., Flocculation of Bulked Activated Sludge
with Polyelectrolytes. Journal Water Pollution Control Federation, 40, No. 2, Part 2,
pp. R1-R9 (1968).
44. Derrington, R.E., Stevens, D.H., and Laughlin, J.E., Enhancing Trickling Filter Plant
Performance by Chemical Precipitation. U. S. EPA, Project 11010 EGL (August,
1973).
45. Brown, J.C., Little, L.W., Francisco, D.E., and Lamb, J.C., Methods for Improvement
of Trickling Filter Plant Performance, Part II, Alum Treatment Studies. U. S. EPA,
Contract No. 14-12-505, University of North Carolina, Chapel Hill, N.C. (1974).
46. Directo, L.S., Miele, R.P., and Masse, A.N., Phosphate Removal by Mineral Addition to
Secondary and Tertiary Treatment Systems. 27th Industrial Waste Conference, Purdue
University, Lafayette, Indiana (May 2-4, 1972).
47. Barth, E.F., and Ettinger, M.B., Mineral Controlled Phosphorus Removal in the
Activated Sludge Process. Journal Water Pollution Control Federation, 39, No. 8,
pp. 1,362-1,368 (August, 1967).
48. Berg, E.L., Brunner, C.A., and William, R.T., Single-Stage Lime Clarification of
Secondary Effluent. Water and Wastes Engineering, 7, pp. 42-46 (March, 1970).
6-26
-------
49. Owne, R., Removal of Phosphorus from Sewage Plant Effluent with Lime. Sewage &
Industrial Wastes, 24, No. 5, pp. 548-556 (May, 1953).
50. Lea, W.L., Rohlich, G.A., and Katz, W.J., Removal of Phosphate from Treated Sewage.
Sewage & Industrial Wastes, 26, No. 3, pp. 261-275 (March, 1954).
51. Long, D.A., Nesbitt, J.B., and Kountz, R.R., Soluble Phosphorus Removal in the
Activated Sludge Process. Progress Report for the Water Quality Office, U. S. EPA,
Project No. 17010 EIP, Pennsylvania University, University Park, Pa. (August, 1971).
52. U. S. EPA - District of Columbia Pilot Plant, Monthly Reports, June-October 1972,
Washington, D. C. (1972).
53. Goodman, B.C., and Mikkelson, K.A., Advanced Wastewater Treatment. Chemical
Engineering Desk Book Issue, 77, pp. 75-85 (April 27, 1970).
54. Zenz, D.R. and Pivnicka, J.R., Effective Phosphorus Removal by the Addition of Alum
to the Activated Sludge Process. Proceedings - 24th Industrial Waste Conference,
Purdue University, pp. 273-301 (1969).
55. Anderson, D.T., and Hammer, M.J., Effect of Alum Addition on Activated Sludge
Biota. Water & Sewage Works, 120, No. 1, pp. 63-67 (January, 1973).
6-27
-------
-------
CHAPTER 7
EFFLUENT POLISHING TECHNIQUES
7.1 General
The use of effluent polishing techniques is receiving increasing attention as a practical and
economical method of upgrading existing secondary treatment facilities to obtain improved
organic and SS removal. Effluent polishing is particularly applicable where it is necessary to
increase overall efficiency by 10 to 20 percent.
Four effluent polishing processes are considered in this manual: (1) polishing lagoons, (2)
microscreening, (3) filtration, and (4) activated carbon absorption. The reader is referred to
the Process Design Manual for Suspended Solids Removal (1) for an in-depth discussion of
microscreening and filtration, and to the Process Design Manual for Carbon Adsorption (2)
for detailed design information on carbon columns.
7.2 Polishing Lagoons
Polishing lagoons offer an opportunity for increased organic and SS removal at a minimum
cost. Both aerobic and facultative lagoons can be used for this purpose.
7.2.1 Aerobic Lagoons
Aerobic lagoons are generally subdivided into two groups:
1. Shallow lagoons, with depths ranging from 2.5 to 4.0 feet.
2. Deep lagoons, 7 to 10 feet deep, with aeration devices included to ensure
maintenance of aerobic conditions.
The shallow aerobic lagoon is one in which the algae-bacterial symbiotic interrelationship is
optimized by providing as much light penetration as possible, and by maximizing
photosynthetic efficiency and bacterial oxidation of organic wastes. Operational data from
several shallow aerobic lagoons are presented in Table 7-1. The data indicate consistent BOD
removals throughout the year, but a marked increase in the concentration of SS in the
effluent during the summer periods, when algal activity is at its peak. The seasonal increase
in residual SS without a concurrent increase in BOD during the summer months is caused by
algae carry-over into the effluent. This indicates that algae present in the effluent do not
exert a significant amount of BOD demand during the five-day incubation used in the
standard BOD test. The substantial increase in effluent SS during the summer period,
however, is a major disadvantage of the shallow lagoon as a dependable year-round polishing
technique.
7-1
-------
TABLE 7-1
OPERATIONAL DATA FROM SHALLOW AEROBIC POLISHING LAGOONS
Plant
Location
Marlborough,
Mass.
Indian Creek,
Kansas
Sedalia,
Missouri
Time Detention
Period
1962
Feb. 13-14
June 13-14
Aug. 30-31
1963
June 12-13
July 10-11
Aug. 27-28
Dec. 1-2
1964
Jan. 28-29
March 6-7
April 10-11
Year of 1969
Time
days
-
-
2.2
2.7
2.4
3.4
3.9
3.0
1.8
-
Flow
mgd
1.4
1.1
0.9
2.3
1.9
2.1
1.5
1.3
1.7
2.8
1.8
BOD
In
Out
mg/1
45
36
26
19
11
11
30
38
27
36
24
15
29
24
10
9
5
15
16
15
11
11
Lagoon
BOD
Removal
percent
67
19
8
50
20
58
50
59
44
69
54
Surface Organic
Loading
Ib BOD/day/acre
47
30
18
61
29
32
61
67
62
135
30
SS
In
Out
mg/1
13
44
39
29
15
11
43
64
20
21
-
9
31
41
52
31
27
17
34
9
5
-
Lagoon
SS
Removal
percent
35
30
-5
-77
-103
-140
61
45
23
77
Remarks
Trickling Filter
Effluent
Trickling Filter
Effluent
Pond Size-6.2 acres
Depth - 2.5 ft
Odorous in spring
Reference
Activated Sludge
Effluent
*See Case History, Chapter 13.
-------
An alternative to the shallow lagoon is the deep, aerated, lagoon. These deeper lagoons can
operate at greater surface organic loadings than shallow lagoons and yet maintain higher
organic removals. Settling of the lagoon effluent is normally necessary to provide efficient
SS capture. This can be provided by a quiescent zone at the end of the lagoon, or by a
separate clarifies Since oxygen is supplied to the basin by mechanical devices rather than
furnished by the algae-bacterial biosymbiotic relationship, the algae production in the
aerated lagoon is minimal compared to the shallow lagoon. Operational data for aerated
effluent polishing lagoons (5 to 10 feet deep) are presented in Table 7-2.
TABLE 7-2
REMOVAL EFFICIENCIES FOR DEEP AERATED
EFFLUENT POLISHING LAGOONS
Surface
Organic BOD SS
Plant Location Loading Removal Removal Reference
Ib BOD/day/acre percent percent
Washington Borough, N. J. 1 230 63 78 5
East Windsor Township, N.J. 2 134 75 75 ^
Indian Creek, Kansas3 60 to 178 60 50 4
1 Low-Rate Trickling Filter Plant.
^Contact Stabilization Plant.
3High-Rate Trickling Filter Plant.
The lagoon at the Washington Borough Plant has average influent BOD and SS
concentrations of 43 mg/1 and 70 mg/1, respectively. The average effluent BOD and SS
concentrations are 16 mg/1 and 15 mg/1, respectively. The East Windsor Plant's polishing
lagoon receives BOD and SS concentrations as high as 80 mg/1, while the effluent
concentrations are generally about 15 mg/1.
In 1965, the Indian Creek Plant capacity was increased to serve a population of 70,000.
Under these conditions it was estimated that the BOD loading to the shallow aerobic lagoon,
previously discussed, would increase to 450 Ib/day/acre, causing a deterioration of the
effluent quality shown in Table 7-1. The shallow lagoon was modified by raising the berm
elevation, installing a new effluent structure, deepening the lagoon to 5 feet and providing a
perforated hose aeration system. After the above modifications, the lagoon received a flow
of 4.0 to 7.0 mgd and provided a detention time of 1.4 to 2.4 days. During 1969, effluent
7-3
-------
BOD and SS concentrations averaged 10 rag/1 and 9 mg/1, respectively. The highest monthly
average effluent BOD and SS concentrations during this time were 17 mg/1 and 14 mg/1,
respectively.
Deep aerated polishing lagoons may be aerated by either mechanical or diffused air systems.
These devices must be designed to provide sufficient oxygen for biological metabolism and
adequate mixing. Mechanical surface aerators are commonly used for this purpose in the
Southern and temperate climate areas, while diffused air systems have been successful in
severe climates where icing may prove troublesome for the mechanical devices.
Eckenfelder (6) has indicated that the power levels required to maintain solids under
suspension and to disperse oxygen uniformly throughout the basin are 0.02 to
0.03 hp/1,000 gallons and 0.006 - 0.01 hp/1,000 gallons, respectively.
Edde (7) studied the degree of mixing provided by mechanical aerators used in treating
wastewater from pulp mills. His study indicates that a velocity greater than 0.4 fps should
be maintained in the basin to prevent solids deposition, and that mixing energy input varies
with the size of the aeration unit. The following values were given as sufficient mixing
energy to disperse oxygen uniformly throughout the basin (7):
TABLE 7-3
MECHANICAL MIXING ENERGY REQUIRED FOR
OXYGEN DISPERSION
Size of Aerators Mixing Energy
hp hp/1,000 gal
100 0.014
50 0.018
20 0.021
The above discussions indicate that mechanical aerators can be designed to provide either
complete mixing of solids including oxygen dispersion, or just to provide uniformly
dispersed oxygen. In the latter case, solids deposition will occur in the basin.
7.2.2 Facultative Lagoons
Facultative lagoons are characterized by two distinct zones - aerobic and anaerobic.
Hydraulic and organic loadings are such that the DO in the lower section of the lagoon is
depleted but an aerobic layer is maintained near the surface. A cross section of a typical
facultative lagoon is shown on Figure 7-1.
7-4
-------
FIGURE 7-1
TYPICAL CROSS SECTION OF A FACULTATIVE LAGOON
71
on
2-3 FT AEROBIC ZONE
TRANSITION ZONE
ANAEROBIC ZONE >3 FT
-------
At Peoria, Illinois, Fall (8) investigated the efficiency of a 10-foot deep polishing lagoon
operated for 9-month periods each as a facultative lagoon and as an aerated lagoon. The
results of his work are summarized in Table 7-4. It is interesting to note that both the BOD
and the SS concentrations in the effluent did not change appreciably during the period
when the lagoon was operated aerobically as compared to the facultative operation. Fall also
has stated that during the two winters of operation there was no ice on the pond. The
lowest temperature of the pond effluent was 48 deg F, and this was recorded after five days
during which air temperatures were below zero degrees F. Facultative operation of the
lagoon produced small amounts of algae in the pond during the summer period, but no odor
problems were noted during the operation of this lagoon.
TABLE 7-4
COMPARISON OF OPERATIONAL DATA FROM FACULTATIVE AND
AERATED POLISHING LAGOONS
Description
Type of Secondary Plant
Flow, mgd
Lagoon Size, acres
Average Pond Depth, feet
Influent BOD, mg/1
Effluent BOD, mg/1
BOD Removal, percent
Influent SS, mg/1
Effluent SS, mg/1
SS Removal, percent
Detention Time, days
Organic Surface Loading,
Ib BOD/day/acre
Air Applied, cu ft air/lb
BOD applied
Odor
Minimum Temperature of
Lagoon During Study, deg F
Sources: Peoria - Fall (8)
Decatur - Reynolds
Aerated
Peoria,
Illinois
Facultative
Activated Activated
Sludge
0.66
0.45
10
58
34
41
55
17
67
1.8
710
223
None
48
(9)
Sludge
0.65
0.45
10
62
30
52
55
18
67
1.82
747
0
None
48
Springfield - Hickman (10)
Decatur, Springfield,
Illinois Missouri
Facultative Facultative
Trickling
Filter
6.8
8.4
5.5
30
18
40
61
31
49
2.3
218
0
None
52
Activated
Sludge
18.7
10
12
83
30
64
69
26
62
1.61
1,292
0
7-6
-------
Operational data from the facultative effluent polishing lagoon in Decatur, Illinois, also
shown in Table 7-4, indicate that BOD and SS removals averaged 40 percent and 49 percent,
respectively, while operating under an organic surface loading of 218 Ib BOD/day/acre (9).
As seen in Table 7-4, the facultative lagoon at Springfield, Missouri, receives much higher
surface BOD loadings (approximately 1,290 Ib BOD/day/acre) and still performs creditably,
with average BOD and SS removals of 64 and 62 percent, respectively (10).
A major disadvantage of facultative lagoons is the fact that the effluent will have a minimal
DO content. Springfield, Missouri, solved this problem by using cascade aeration. The
effluent from the polishing lagoon flows over a series of five weirs with a total drop of
75 inches. The average DO in the effluent (September 1970 through March 1971) was
7.0 mg/1 with a minimum and maximum, respectively, of 4.0 and 9.9 mg/1 (10).
7.3 Microscreening
7.3.1 Principles of Operation
The microscreen is a surface filtration device that has found increasing utility for polishing
secondary effluents. It consists of a specially woven polyester or stainless-steel screen
mounted on the periphery of a partially submerged, horizontal revolving drum. Influent
wastewater enters the drum internally and passes radially outward through the screen, with
deposition of solids on the inner surface of the drum screen. A typical microscreen unit is
shown on Figure 7-2.
Microscreens are continuously backwashed by water jets located at the top of the drum.
These jets normally operate at a pressure of 15 to 50 psig. The backwash water is returned
to the head of the plant, and usually totals 4 to 6 percent of the microscreen throughput.
The screens employed have extremely small openings and are available in a variety of sizes,
as shown in Table 7-5.
TABLE 7-5
MICROSCREEN FABRIC SIZES
Number of
Opening Openings
y per sq in.
15
23 165,000
35 80,000
60 60,000
7-7
-------
8-Z
I(Nn
-------
Typical microscreen drum sizes, capacities, power and space requirements are shown in
Table 7-6.
TABLE 7-6
TYPICAL MICROSCREEN POWER AND SPACE REQUIREMENTS
Approximate
Floor Space
Diameter Length Width Length
Drum Sizes
ft
ft
ft
ft
5.0
5.0
7.5
10.0
1.0
3.0
5.0
10.0
8
9
11
14
6
14
16
22
0.50
0.75
2.00
5.00
Motors
Drive
BHP
0.50
0.75
2.00
5.00
Wash Pump
BHP
1.0
3.0
5.0
7.5
Approximate
Ranges of
Capacity
mgd
0.07 - 0.15
0.2 -0.4
0.5 - 1.0
1.5 -3.0
Courtesy Crane Company
4.0
6.0
10.0
4.0
6.0
10.0
7
10
14
15
17
22
0.75
2.00
5.00
1.0
1.5
5.0
0.2 -0.4
0.5 - 1.0
1.5 -3.0
Courtesy Zurn Industries
The weave and shape of individual fabric wires are such that they allow the water from the
backwashing jets to penetrate the screen and remove the solids mat which forms on the
inside of the screen during its passage through the feed stream. Bodien and Stenburg (11)
have noted that only about one-half of the applied washwater actually penetrates the screen;
the rest flows down the outer perimeter into the effluent chamber. Previously screened
effluent can be used as washwater.
As a section of the screen passes through its cycle, the resistance to flow increases as the
solid mat forms. Standard design calls for a three-inch head loss at average flow and a
six-inch head loss for maximum flows. Newer microscreen models are available with
automatic controls to increase drum speed and backwash pressure to accommodate variation
in flow and, to a lesser extent, variation in solids loading. Ultraviolet light placed in close
proximity to the screen has been somewhat successful in slowing the development of screen
clogging slimes. In general, however, the units must be taken out of service on a regular basis
(once a month, for example) to have the screens cleaned. In addition, cleaning may be
required for removal of iron, manganese, or grease. In cases where oil and grease problems
occur, a hot water or steam treatment can be used to remove these materials from the
screen.
7-9
-------
One of the advantages of using a microscreen for upgrading the performance of existing
plants is its low head requirement. It is therefore advantageous to transfer secondary
effluent without pumping, to a tertiary microscreening installation to minimize the shear
force imparted to the relatively fragile biological floe. Careful design of inlet structures and
outlet structures will eliminate turbulent areas and hold the head loss through the entire
installation to 12 to 18 inches (12). Typical designs include overflow weirs to bypass part of
the flow when the head loss through the screen exceeds a predetermined amount, usually six
to eight inches.
7.3.2 Functional Design
Functional design of a microscreen unit involves:
1. Characterization of SS in feed as to concentration and degree of flocculation. The
character of the solids affects microscreen capacity, performance and
backwashing requirements.
2. Selection of unit design parameter values which will: a) assure capacity to meet
maximum hydraulic loadings with critical solids characteristics, b) provide desired
performance over the expected range of hydraulic loadings and solids
characteristics.
3. Provision of backwash and supplemental cleaning facilities to maintain capacity.
Table 7-7 shows typical microscreen and backwash design parameter values for tertiary
solids removal applications.
7.3.3 Performance
Suitable relationships have not been developed for quantitative predictions of microscreen
performance from knowledge of influent characteristics and key design parameters.
Where performance must be predicted closely, pilot studies should be made. Where close
prediction is less critical, performance data from other locations with generally similar
conditions may serve as a guide.
Microscreening has been used for the removal of algae from lagoon effluents. At Bristol,
England, algae reductions of 1,565 to 450 algae per ml and 989 to 168 algae per ml were
achieved on astrerionella, cyclotella and synedra (1).
Many classes of algae (e.g., chlorella) are, however, too small to be removed, even on fine
screens (23 M )and excessive loadings (up to 2 x 10^ algae per ml) make this application a
limited one.
7-10
-------
TABLE 7-7
TYPICAL MICROSCREEN DESIGN PARAMETERS
Item
Screen Mesh
Submergence
Hydraulic Loading
Head Loss (H.L.)
through Screen
Peripheral Drum Speed
Diameter of Drum
Backwash Flow and
Pressure
Typical Value
20-25 u
75% of height
66% of area
5-10 gpm/sq ft of sub-
merged drum surface
area
3-6 in.
15 fpm at 3 in. H.L.
125-150 fpm at 6 in.
H.L.
10ft
2% of throughput at
50 psi
5% of throughput at
15 psi
Remarks
Range 15-60
Maximum under extreme condition:
12-18 in. Typical designs provide for
overflow weirs to bypass part of flow
when head exceeds 6-8 in.
Speed varied to control H.L. Extreme
maximum speed 150 fpm
Use of smaller diameters increases
backwash requirements
7-11
-------
Table 7-8 provides performance data for a number of microscreen applications for tertiary
solids removal. Reported data indicate the following:
1. Even under best operation, 5 to 10 mg/1 residual SS will pass through the
microscreen unit.
2. Although the pattern is irregular, performance tends to be better at lower
hydraulic loadings.
3. Increases in influent SS are reflected in the effluent but with noticeable damping
of peaks.
4. Better removals are obtained with smaller mesh size.
7.4 Filtration
7.4.1 General
Filtration of secondary effluent provides a positive method of SS control, and as such, is
one of the more widely used and the most efficient single unit process for upgrading
treatment plant performance today. The primary applications of filtration are: (1) direct
filtration of secondary effluent, (2) filtration of chemically clarified secondary effluent, and
(3) filtration of secondary effluent after in-line chemical injection. Direct filtration has the
highest probability of use where existing plants will be required to consistently meet better
effluent quality standards than they are presently able to attain.
The major goals of filtration design are:
1. Attainment of required effluent quality
2. Low capital cost
3. Low operating cost.
Present filtration technology provides the designer with a broad selection of filter types and
sizes, all of which have been shown capable of producing high quality effluents. For any
particular application, adequate consideration must be given to matching filter design and
capability to the existing site conditions and wastewater characteristics. Such factors as
present and projected plant flows, variability and concentration of filter influent SS, plant
hydraulics, and present and projected operation schedules all affect filter selection (13).
7-12
-------
TABLE 7-8
MICROSCREEN PERFORMANCE DATA (1)
Location
Harpenden, England
Luton, England
Brackncl], England
G. L. C Rcdbridgc, England
Hamblrdon R.D.C.
Elmbridgc, England
Leighton-Lmslade U.D.C.
England
Fleet U.D.C., England
Eshcr U.D.C. , England
Hatfield R.D.C.. England
The Borough of Bury
St. Edmonds, England
Essex R.B. Works, England
Franklin Township
STP Murraysville, Pa.
Letchworth, England
Bahinstokc, England
Euclid, Ohio
Euthd, Ohio
Euclid, Ohio
Lebanon, Ohio
Hanover Park, III.
MSD North Side STP
Chicago, 111.
Miami, Fla.
Murfreesboro, Tenn.
Essex Junction, Vermont
Source of Micro-
screener Influent
Stream
Trickling Filters
Humus-Tank *
Humus-Tank
Humus-Tank
Humus Tank
Humus-Tank
Humus-Tank
Humus-Tank
Humus-Tank
Humus-Tank
Humus-Tank
Trickling Filters
Final Clanficrs
Activated Sludge
Final Clarifiers
Activated Sludge
Final Clarifiers
Activated Sludge
with Chem Precip.
in Primary Clarif.
(FeCl)
Activated Sludge with
Chem. Precip. in Final
Clarif. (FeCl)
Pure Oxygen-Activated
Sludge Final Clanficrs
Activated Sludge
Final Canf.
Attivaled Sludge
Final Clarif.
Activated Sludge
Final Clarif.
Drum
Diameter
Width
ft
5x3
-
7.5x5
-
-
-
-
10x10
lOx 10
7.5 x 5
lOx 10
-
lOx 10
5x3
lOx 10
2.5x2
2.5 x 2
2.5x2
Sx 1
5x 1
lOx 10
12.5x30
lOx 10
lOx 10
4x4
Screen
Mesh
V
'65
60
35
23
23
35
35
35
-
-
23
-
23
23
23
23
23
23
35
23
23
23
20
20
35
Hydraulic
Load on
Submerged
No. of Area
Units Max.
gpm
1 6.6
9.0
2 6.3
9.4
10.8
10.8
68
2 6.0
3
3
5
9.8
2 7.8
1 4.3
5 1.5
1 2.5
1 2.5
1 2.5
1
1
Avg.
/sqft
Plant Flow
Max.
mgi
Avg-
il
Pilot Study
-
2.2
6.8
2.5
2.5
3.9
2.0
-
-
-
6.8
-
3.3
1.0
1.25
1.25
1 25
-
-
1 5 3 2.6
-
1
2
1
-
-
_
-
3
6.3
2.7
2.5
25
2.0
3.6
10.82
2.552
9.02
5.3
2.8
4.0
-
2.2
2.0
0.6
0.6
1.2
1.2
-
2.0
-
Pilot Study
3.2
40
gpm
40
gpm
40
gpm
-
1 5
152
2.7
4.0
0.25
2.2
20
gpm
20
gpm
20
gpm
1.0
08
-
-
-
Influent
mg/1
40
14
20
16
30
14
14
29
15
19
14
28
26
37
17
13
54
38
65
27
17
6-28
10
-
_
-
Average SS
Effluent
mg/1
11
8
11
7
15
8
8
11
6
9
8
7
14
6
6.6
4
8
10
21
7
2
4-11
3
-
_
-
Removal
percent
73
45
45
57
50
45
45
60
60
60
43
75
44
83
62
70
85
74
68
73
83
55 (avg.)
67
71
50
50
Manufacturer
Crane
Crane
Crane
Crane
Crane
Crane
Crane
Crane
Crane
Crane
Crane
Crane
Crane
Crane
Crane
Crane
Crane
Crane'
Crane
Crane
Zurn
Zurn
Zurn
I. Trickling Filters Final Clanfier.
2 Design Flow.
7-13
-------
Low capital costs will generally result from low filter surface areas, and high filtration rates.
Operating costs are reduced by designs that minimize backwashing requirements and
maximize filter run lengths. In wastewater filtration, run lengths are terminated by either
excessive head loss or by deterioration of effluent quality beyond allowable limits. The
optimum bed design is one that reaches these limiting conditions at the same time. In some
cases, especially in small plants, it may be desirable to design the filter beds so that routine
backwashing can be performed once a day.
7.4.2 Filter Types
Filtration systems can be broadly classified as either in-depth or surface. In-depth systems
include deep-bed single coarse medium filters as well as dual- or tri-media filters. The use of
two or more layers of different media having increasing specific gravity in the direction of
flow allows gradation of the filter bed from coarse to fine. This allows more efficient
utilization of the total bed depth for solids storage than conventionally graded,
single-medium filters. Multimedia beds normally require more backwash water and higher
backwash rates (25-30 gpm/sq ft) than single-medium beds (13). Typical multimedia
gradations are shown in Table 7-9. The use of several media such as garnet, sand and coal
will allow some flexibility in bed design to meet specific effluent quality or desired run
lengths.
TABLE 7-9
TYPICAL MULTIMEDIA GRADATIONS (1)
Garnet Sand Coal
Gradation Size Depth Size Depth Size Depth
No. (Mesh) (Inches) (Mesh) (Inches) (Mesh) (Inches)
1 -40x80 8 -20x40 12 -10x20 22
2 -20 x 40 3 -10 x 20 12 -10 x 16 15
3 -40x80 3 -20x40 9 -10x20 8
The approach taken in the development of surface filters is to allow filtration to take place
on or near the top of relatively shallow (12 in.) single medium filters, and to optimize
removal of the accumulated solids. In addition to the standard pressure and gravity surface
filters, several innovative techniques for maintaining a continuous clean filter surface have
been developed. These include moving bed filters, radial flow filters, radial-flow external
wash filters, and traveling-bed filters.
7-14
-------
Although both upflow and downflow modes of filter operation are in use, the downflow
filter is far more common for wastewater applications.
As a rule, in-depth filters are better suited to treating strong biological floe, will yield longer
run lengths, and are less sensitive to solids loading than are surface filters. Surface filters are
better adapted to removing more fragile floes, yield shorter run lengths, and require less
backwash water per cycle than in-depth filters.
7.4.3 General Design Considerations
7.4.3.1 Filter Selection and Governing Factors
Performance of different filter systems on a given wastewater may be termed "equivalent" if
they produce the same output quality and quantity (1). The major factors that determine
overall filter performance are summarized in Table 7-10.
TABLE 7-10
FACTORS GOVERNING FILTER PERFORMANCE
1. Maximum Available Head Loss
2. Filtration Rate
3. Influent Characteristics
a. SS concentration
b. Particle size distribution
c. Floe strength
d. Temperature (viscosity)
e. Properties governing adhesion of solids to each other or to media
4. Media Characteristics
a. Grain size
b. Porosity
c. Depth
d. Specific gravity
e. Configuration
5. Design of Cleaning System
a. Adequacy of cleaning
b. Washwater volume per cleaning cycle
7-15
-------
In general, variations in these factors which increase quality, such as use of finer media, tend
to increase head loss and hence reduce output whereas those which increase output tend to
reduce quality. Run length may be limited by available head or effluent quality
(breakthrough). Washwater requirements depend chiefly on the size gradation, specific
gravity of the media, and on the type of supplementary cleaning available.
Another factor that must be examined when selecting a filter for a given wastewater is the
effect of chemical pretreatment. Typical situations include activated sludge and trickling
filter plants that incorporate alum or iron addition followed by filtration for high level
phosphorus removal. In these cases the filter performance may be quite different than for
untreated effluents due to the weaker chemical floe normally produced. Generally it is
desirable to provide for the addition of polyelectrolytes in installations where aluminum or
iron salts are employed for phosphorus removal.
Since influent solids characteristics are variable, the operating head provided in design must
cover a range of conditions. If breakthrough consistently requires termination of filter runs
at a low head loss, much of the available head will be wasted. On the other hand, if available
head is so low that head loss instead of quality limits the run length, filter production
capacity will be impaired unnecessarily.
Based on long term practice in water filtration, maximum terminal head loss of 10 to
15 feet has commonly been allowed. In water filtration, this head loss range generally
permits runs of 24 or even 48 hours. In wastewater applications, however, heavy solids
loadings and high biological floe strength can cause rapid head loss buildup, frequently
reaching limiting values well below 24 hours. Filter design should allow sufficient head so
that, with maximum influent solids concentrations and floe strengths, minimum run lengths
of six hours can be achieved. Also the effect of backwash recycle flows and filter downtime
must be considered when sizing units.
Proper cleaning is vital to filter performance. Ineffective cleaning results directly in short
filter runs and poor effluent quality and if continued, sets up self-perpetuating operational
difficulties such as mud balls, slime coating of the media and cracking of the filter bed (14).
In wastewater filtration, normal upflow washing plus auxiliary cleaning (air wash or surface
wash by water jets) and periodic shock chlorination have proven necessary to maintain
filters in proper condition. During a normal backwashing operation, accumulated solids are
removed from filters by a rapid upflow of washwater which is returned to the previous
treatment units. In small plants, equalization of this flow is necessary to prevent surcharge
of upstream treatment units. Washwater sources may include feedwater, filter effluent, or
effluent from subsequent treatment units. Washwater storage may be needed if rates
required exceed effluent flow available. Backwash rates for most effective cleaning vary with
media size, particle density, floe penetration, and strength of floe adherence to the media.
7-16
-------
In most applications a number of specific filter designs can provide the required
performance. For a given treatment application, if costs and performance variations of
alternatives can be defined quantitatively, it is possible to base selection on "least cost."
Cost comparisons should include both installed cost and operating cost items such as control
required, operating head loss, and backwash water consumption.
7.4.3.2 Pilot Plant Investigations
The extent of pilot investigations which should be conducted as a basis for a given design
depends on the size of the proposed installation and on the performance requirements
which must be met. For large installations the cost of the necessary pilot plant investigations
for optimizing filter design is certainly justified in view of the potential cost savings. When
undertaken, these studies should be sufficiently well planned to relate filter performance to
the full range of variables expected under actual operating conditions. Figure 7-3 illustrates
the type of information that should be developed as a part of an overall pilot investigation
to optimize filter design. This figure represents the results of pilot plant studies including
only those runs that produced effluents of satisfactory quality. From such data a rational
design can be formulated.
Where quality requirements are less stringent, or for very small plants where extensive
efforts at cost optimization are not justified, filter design parameters may be selected based
upon evaluations of similar installations.
7.4.4 Performance
The single most important factor affecting filter performance is the quality of the secondary
effluent produced by the biological treatment. If consistently good performance is exhibited
by the biological treatment system, good filter performance can be expected. Conversely, if
the biological facility is subject to frequent upsets, filtration will be much more difficult.
Data are presented in Tables 7-11 and 7-12 relating expected filter performance with degree
of treatment in activated sludge and trickling filter plants. The data in the tables are based
on a filter rate of 5 gpm/sq ft, using a 30-inch deep multimedia filter (55 percent coal,
30 percent sand, and 15 percent garnet) operated to a 10-foot head differential.
Pilot and full-scale performance data for in-depth filters operating on activated sludge and
trickling filter plant effluents are shown in Table 7-13. Performance data for surface filters
are shown in Table 7-14. As indicated in these tables, both in-depth and surface filters
produce an effluent with BOD and SS concentrations generally less than 7 mg/1.
Data describing filter performance at various plants treating chemically treated secondary
effluent are shown in Table 7-15.
7-17
-------
60
FIGURE 7-3
TYPICAL PILOT PLANT DATA FOR FILTER DESIGN (13)
I r
50
oo
CO
40
30
2 20
2 GPM/SQ FT
6 GPM/SQ FT
I
I L
10 20 30 40
INFLUENT SOLIDS CONCENTRATION, MG/L
50
7-18
-------
TABLE 7-11
EXPECTED FILTER PERFORMANCE FOR
ACTIVATED SLUDGE PLANTS (15)
Good Biological Treatment
Type of Activated
Sludge Process
Conventional and
Extended Aeration
Contact Stabilization
Type of Activated
Sludge Process
Conventional and
Extended Aeration
Contact Stabilization
Filter Influent Filter Effluent
BOD SS BOD SS
mg/1 mg/1 mg/1 mg/1
12-15 15-25 2-5 1-4
15-20 15-25 5-10 1-5
Fair Biological Treatment
Filter Influent Filter Effluent
BOD SS BOD SS
mg/1 mg/l mg/l mg/l
20-35 30-50 5-10 5-10
30-45 25-50 20-25 5-10
TABLE 7-12
Run Time
hr
16-24
12-20
Run Time
hr
6-12
6-10
EXPECTED FILTER PERFORMANCE FOR
Percent
85 Percent
Filter Filter
Influent Effluent
BOD SS BOD SS
me/1 mg/1 mg/1 mg/1
TRICKLING FILTER PLANTS (15)
Soluble BOD Removed in Secondary Process
80 Percent
Filter Filter
Influent Effluent
RunTime BOD SS BOD SS
hr mg/1 mg/1 mg/1 mg/1
Run Time
hr
30-40 30-40 20-30 15-20
6-11
40-50 35-45 30-40 20-25
5-9
7-19
-------
TABLE 7-13
IN-DEPTH FILTRATION OF ACTIVATED SLUDGE
AND TRICKLING FILTER PLANT EFFLUENTS
Location
Goldwater, Michigan
Bedford Twp., Michigan
Walled Lake-Novi,
Michigan
Pontiac, Michigan
State College, Pa.
(Spring Creek)
Louisville, Kentucky
(Hite Creek)
Ann Arbor, Michigan
Philomath, Oregon
Ventura, California
Hanover Park, Illinois
Note. () indicated range
Influent
Source
Trickling
Filter
Activated
Sludge
Activated
Sludge
-
Activated
Sludge
Activated
Sludge
Activated
Sludge
Extended
Aeration +
Tube
Settler
Trickling
Filter
Activated
Sludge
Activated
Sludge
Activated
Sludge
Activated
Sludge
Activated
Sludge
of values.
Type of Filter
Horizontal
Pressure
Horizontal
Pressure
Gravity
Downflow
-
-
Pressure
-
Pressure
Downflow
Gravity Deep
Bed Downflow
Pressure
Dpflow
Pressure
llpflow
Pressure
Upflow
-
-
Hydraulic
Media Loading
gpm/sq ft
Coal-(18 in )
Sand-(12 in.)
Garnet
Multimedia
Multimedia 3 to 4
Multimedia
-
Multimedia 3.4
Multimedia 6
Multimedia 5
1.2mm 6
Sand
Multimedia 2.2
Multimedia 4.0
Multimedia 4.9
Multimedia 2.0
Multimedia 4.0
Filter Influent
BOD
mg/1
(32-56)
44
(15-20)
-
-
9
12
(3-19)
17
(11-50)
22
(7-36)
26
23
17
24
20
23
13
SS
mg/1
(13-29)
21
(10-25)
15
(5-10)
7
19
12
(11-32)
27
(28-126)
42
(30-2180)
165
(19-21)
18
14
15
13
14
16
Filter Effluent
BOD
mg/1
(16-42)
30
(4-10)
7
-
2
3
2
(3-8)
5
(1-4)
3
18
6
6
7
6
2
SS
mg/1
(4-12)
8
(2-5)
3
(1-4)
3
2
4
(1-4)
3
(1-17)
5
(1-20)
5
7
7
5
6
4
4
Length of
Filter Runs Reference
hr
2.5 to 8 16
15 16
16
17
12 18
19
17
15-24 20
6-18 21
22
, 22
22
22
22
7-20
-------
TABLE 7-14
SURFACE FILTRATION OF ACTIVATED SLUDGE
AND TRICKLING FILTER PLANT EFFLUENTS
10
Influent Source
Activated Sludge
Activated Sludge
Activated Sludge
Activated Sludge
Activated Sludge
Activated Sludge
Contact Stabilization
Trickling Filter
Filter
Type of Filter
Gravity Downflow
Gravity Downflow
Gravity Downflow
Gravity Downflow
Gravity Downflow
Gravity Downflow
Gravity Downflow
Gravity Downflow
Media
Coal - 30 in.
Sand - 12 in.
Garnet - 6 in.
Coal - 30 in.
Sand - 12 in.
Garnet - 6 in.
Coal - 30 in.
Sand - 12 in.
Garnet - 6 in.
Hydraulic
Loading
gpm/sq ft
2.2
4.0
8.0
1.6-4.0
2.0
2.0-6.0
5.3
0.75
Influent
BOD
mg/1
20
25
19
18
47
SS
mg/1
16
15
12
14
36
Filter
Effluent
BOD
mg/1
7
7
7
52-70%*
57%'
80% l
4
17
SS
mg/1
7
5
6
72-91%'
46% i
70% 1
5
10
Reference
' 22
22
22
23
24
25
26
16
Removal Efficiencies.
-------
TABLE 7-15
FILTRATION OF CHEMICALLY TREATED SECONDARY EFFLUENT
Type of Plant
Location Influent Source Filter Capacit\
Piscataway, Maryland Activated Sludge + Pressure 5 mgd Dual
2-Stage Lime Clari- Downflow
fication
Ely, Minnesota High Hate Trick- Gravity 1.5 mgd Dual
ling Filter + Downflow
2-Stage Lime Clari-
fieation
Hydraulic
Loading
gpm/sq ft
3
2.3
3
Length
of
Filter
Run BOD
In mg/l
50
24
2.5-6.5 (30-130)
75
Filter Influent
COD TOC
mg/l mg/l
9
i -j
Filter Effluent
SS
mg/l
12
8
(20-70)
40
BOD COD TOC
mg/l mg/l mg/l
19 8
6-15
(18-110) -
46
SS
mg/l
8
<2
(5-35)
21
Reference
27
28
29
Jefferson Parish, Trickling Filter Upflow 0.5 mgd Sand
Louisiana and In-Line Alum
Injection
Lebanon, Ohio Activated Sludge Pressure 50 gpm Dual 5 2.5-4 - - (10-30) (20-35) - - (5-15) <1 30
with In-Line Alum Downflow
and Polyelectro-
lyte Injection
Pomona, California Activated Sludge Gravity 30 gpm Dual 3 24 81 10 21 - 5 31
with Tertiary Downflow
Alum and Poly-
flcctrol)tc
Clarification
Note ( ) indicates range of values.
-------
While it is recognized that filtration efficiency varies depending on the type of filter, bed
design, characteristics of the wastewater, and many other factors previously discussed, it is
informative to examine the limits of this variability. An evaluation of filtration data from a
number of installations indicates that removal of SS averages approximately 70 percent,
with a range of 50 to 90 percent.
7.5 Activated Carbon Adsorption
The limitations of conventional biological treatment processes in regard to reliable
achievement of a high degree of organic removal (particularly of certain compounds which
are refractory to biodegradation), along with increasingly strict water quality standards,
emphasize the need for a supplementary organic removal process. Thus, activated carbon is
presently being used to provide tertiary treatment of biologically treated effluents.
Experience gained from the operation of activated carbon plants for tertiary treatment of
wastewater suggests that activated carbon need not be restricted to a polishing role, but can
be used as an alternative to biological treatment. Replacement of conventional biological
treatment by activated carbon (i.e. secondary treatment application) is emphasized in the
Process Design Manual for Carbon Adsorption (2). The following discussion is concerned
exclusively with the tertiary application of carbon.
Activated carbon for wastewater treatment can be used either in the powdered or in the
granular forms. The impracticality of economical regeneration has restricted the use of
powdered carbon in wastewater treatment, although this problem is being resolved.
7.5.1 Process Principles and Design Factors
The adsorption of organic materials from wastewater onto the activated carbon involves
complex physical and chemical interactions. Biological degradation of adsorbed materials
also occurs, and this can significantly enhance the overall treatment performance (32) (33).
The ability of activated carbon to adsorb large quantities of dissolved materials from
wastewater is due to its highly porous structure and to the resulting large surface area, which
provides many sites for adsorption of dissolved materials.
Important factors in the design of activated carbon treatment facilities
include: pretreatment requirements; particle size; hydraulic loading and contact time;
regeneration losses; flow configuration; and required effluent quality.
7.5.1.1 Pretreatment Requirements
Granular carbon can be used as a direct polishing technique for secondary effluents or may
be preceded by a SS removal process or other treatment processes as required to accomplish
treatment objectives. Long term studies at Pomona, California, have indicated that carbon
7-23
-------
can function as an effective filter in addition to removing dissolved organic matter (34).
However, it is important that the secondary treatment plant clarification system produce a
reasonably high quality effluent (<20 mg/1 SS), since the use of carbon as a media to
remove large quantities of suspended material is not a cost-effective treatment technique.
7.5.1.2 Particle Size
Theoretically, carbon particle size primarily affects the rate of adsorption and not the
capacity of the carbon. Adsorption rates are greater for smaller particle sizes than for larger
particle sizes. However, adsorbents close to saturation will be less affected by particle size
than adsorbents in their virgin state (32).
Data from Lake Tahoe indicates that there will be a reduction in the adsorption capacity of
about 20 to 35 percent in going from 12 by 40 mesh carbon to 8 by 30 mesh carbon at a
relatively short contact time (32). This apparent difference in adsorption capacity
attributable to particle size is minimized at longer contact times (35). Since finer particle
sizes are susceptible to greater head losses, 12 by 40 mesh carbon is probably not suitable
for use in downflow columns (35).
7.5.1.3 Hydraulic Loading Rate and Contact Time
Contact time, hydraulic loadings, and bed depth are interrelated physical parameters. Of the
three, contact time is clearly the most important. Since the activated carbon treatment of
wastewater requires that a definite contact time be established to complete the adsorption
process, any increase in applied hydraulic load necessitates a deeper carbon column to
maintain the same contact time.
Data obtained at the Pomona, California Pilot Plant indicate that Total Organic Carbon
(TOG) removal does .not vary significantly after 15 minutes contact time for hydraulic
loading rates of 4, 7, and 10 gpm/sq ft (32). It was further noted that for equivalent contact
times, the percent TOG removal was similar for hydraulic loading rates of 4, 7 and 10
gpm/sq ft. These results indicate that contact time is more important than applied hydraulic
loadings, and is, in fact, the most important design factor in carbon adsorption systems.
Typical hydraulic loading rates and contact times used in various locations are shown in
Table 7-16. It should be noted that both gravity and pressure systems are available. Gravity
flow systems are not likely to be practical at hydraulic loading rates above about 4 gpm/sq
ft.
7-24
-------
Ol
TABLE 7-16
TYPICAL CARBON COLUMN DESIGN DATA
Average No. of Total
Effluent
1.
2.
3.
4.
5.
6.
7.
8.
9.
10.
11.
12.
13.
Site
Arlington, Virginia
Colorado Springs, Colorado
Dallas, Texas
Fairfax County, Virginia
Los Angeles, California
Montgomery County, Maryland
Occoquan, Virginia
Orange County, California
Piscataway, Maryland
St. Charles, Missouri
South Lake Tahoc, California
Windhoek, South Africa
Pomona, California
Plant
Capacity
mgd
30
3
100
36
52
60
18
15
5
5.5
7.5
1.0
0.3
Contactor Contactors
Type in Series
Downflow
Gravity
Downflow
Upflow
Packed
Downflow
Gravity
Downflow
Gravity
Upflow
Packed
Upflow
Packed
Upflow
Packed
Downflow
Pressure
Downflow
Gravity
Upflow
Packed
Downflow
Pressure
Downflow
Pressure
1
2
1
1
2
1
1
1
2
1
1
2
4
Contact
Time1
min
38
30
10
36
50
30
30
30
37
30
17
30
40
Hydraulic
Loading
gpm/sq ft
2.9
5
8
3
4
6.5
5.8
5.8
6.5
3.7
6.2
3.8
7
Carbon
Depth
ft
15
20
10
15
26
26
24
24
32
15
14
15
38
Carbon Requirements
Size (Oxygen Demand) Reference
mesh
8x
8x
8x
8x
8x
8x
8x
8x
8x
8x
8x
12 x
12 x
30
30
30
30
30
30
30
30
30
30
30
40
40
mg/1
BOD
BOD
BOD
BOD
BOD
COD
BOD
COD
BOD
COD
COD
BOD
BOD
COD
COD
COD
< 3 2
< 2 2
<10 2
< 5 (by 1980)
< 3 2
<12 2
< 1 2
< 1 2
<30 2
< 5 2
< 5 2
<30
<10 2
<12 34
* Empty bed (superficial) contact time for average plant flow.
2 Ultimate capacity t>0 mgd.
-------
7.5.1.4 Effect of Regeneration
Activated carbon requires regeneration when its adsorption capacity is exhausted.
Considerable effort has been expended to determine the effect of regeneration on
adsorption capacity of the carbon. However, since few research groups have regeneration
facilities, only limited data are available. Results obtained at Pomona (32) indicate that the
adsorptive capacity decreases by approximately 35 percent after seven regeneration cycles,
as indicated on Figure 7-4. It was also determined that regeneration does not affect the
degree of organic removal in subsequent exhaustion cycles. This loss of capacity is not
necessarily a critical factor, since it is necessary to make up physical losses of carbon after
each regeneration cycle. These losses are caused by several factors: carbon is burned and
lost through the stack as combustion products; or is abraded into dust in the course of
handling. Further "losses" are due to the buildup of inorganic ash in the carbon particles
during repeated use and regeneration.
7.5.1.5 Flow Configuration
Depending on the dissolved organic and SS loading, any of several optional flow
configurations can be adopted:
1. Downflow Beds in Series the lead contactor is removed', regenerated, and
replaced in line at the downstream end, the other contactors being moved up in
sequence.
2. Downflow Beds in Parallel parallel beds are arranged in a staggered exhaustion
pattern so that when one is exhausted and removed from service, the product of
the others can be blended with that portion of flow normally treated by the
exhausted contractor to maintain the required product quality for the entire
plant.
3. Upflow Beds (expanded or partially expanded) no head loss is built up, and no
backwashing is necessary; postfiltration is required; the same series and parallel
considerations apply as for downflow operation.
4. Upflow (moving bed) exhausted lower strata of the bed are continuously
removed and replaced at the top of the bed by virgin carbon.
Parallel comparison studies between downflow and upflow beds were conducted at the
treatment plant of the Ewing-Lawrence Sewerage Authority located near Trenton, New
Jersey (36). These studies have indicated that upflow and downflow beds have equivalent
adsorption capacities since each system removed approximately the same quantity of
soluble organic material from the plant's secondary effluent (see Table 7-17). The downflow
bed system is more effective for removal of SS, but at the expense of an increased head loss,
7-26
-------
FIGURE 7-4
EFFECT OF REACTIVATION ON ADSORPTION CAPACITY
CJ
CD
50
45 -
CJ
CJ
35 -i
30 -
25
2345
NUMBER OF REGENERATIONS
6 7
7-27
-------
and a corresponding increase in the frequency of backwashing. Accordingly, an upflow bed
system can process more effluent than a downflow bed system of equivalent size because of
less downtime for the carbon-backwash operation. If no allowance is made for backwashing,
an upflow bed system is capable of processing about 9 percent more effluent than a
downflow system of equivalent size (36).
TABLE 7-17
PERFORMANCE OF UPFLOW BED AND
DOWNFLOW BED ADSORBERS (36)
Description Filtered Secondary Effluent Unfiltered Secondary Effluent
Downflow Upflow Downflow Upflow
TOG Removed, percent 52.6 48.1 57.0 52.0
Soluble Organic Carbon
Removal, percent 42.2 44.7 49.6 45.7
Soluble Organic Carbon
Removed per Ib Active
Carbon, Ib 0.19 0.20 0.23 0.22
7.5.1.6 Performance
In addition to the above design considerations, the question of effluent quality standards
should not be neglected.
The effectiveness of granular activated carbon for upgrading the treatment efficiency is well
established. Full-scaJe operating experiences at Pomona and South Lake Tahoe, California;
Colorado Springs, Colorado; and Piscataway, Maryland, have left little doubt regarding
process efficiency, operating cost and reliability of these systems. The principal design
parameters for these plants and other tertiary granular carbon plants were previously shown
in Table 7-16,
Performance data for operating tertiary carbon facilities are shown in Table 7-18. It is
evident from these data that a very high quality effluent can be produced with the addition
of tertiary carbon treatment.
7.5.2 Laboratory and/or Pilot Plant Investigations
Activated carbon removes dissolved materials from wastewaters by a combination of three
mechanisms: adsorption, filtration, and biological degradation. Therefore, in order to judge
7-28
-------
KI
IS3
BOD, mg/1
COD, mg/1
TOG, mg/1
SS, mg/1
Turbidity, Jtu
Color (Platinum-Cobalt)
Odor
MB AS, mg/1
TABLE 7-18
PERFORMANCE OF TERTIARY CARBON
WASTEWATER TREATMENT PLANTS
Pilot Plant
Pomona, Calif. (34)
Lake Tahoe, Calif. (2)
Piscataway, Md. (27)
Pilot Plant
Orange County, Calif. (2)
Carbon
Influent
3
43
12
9
8
28
12
Effluent
1
10
3
0.6
1
3
1
Secondary
Effluent
30
70
26
15
Carbon »
Influent
3
25
0
0.3
Effluent
0.7
10
0
0.3
Secondary
Effluent
32.9
11.9
15.7
Carbon
Influent
18.6
7.8
7.7
Effluent
6.1
1.5
1.0
Carbon
Influent
30-80
100-200
30-80
Effluent
2
10-30
1
2.0
0.5 0.1
3-4 0-1
-------
the effectiveness of activated carbon for wastewater treatment, both laboratory and pilot
testing are required.
The adsorption mechanism can be evaluated in the laboratory by running adsorption
isotherms. Actual plant conditions should be simulated with regard to temperature, pH, and
pretreatment. A detailed isotherm procedure is given in many books, as well as in the
Process Design Manual for Carbon Adsorption (2).
Adsorption isotherms are normally conducted by contacting a sample of wastewater with
varying amounts of pulverized carbon for a standard interval of time. The wastewater
sample is analyzed for TOG, COD, or BOD (as deemed necessary), both before and after
contacting with the pulverized activated carbon. The treated water should be coarse-filtered
prior to analysis to eliminate carbon fines. The isotherm is a plot of the amount of solute
adsorbed per unit weight of carbon as a function of residual concentration of solute. The
isotherm is empirically represented by the following expression (23):
x/m = KG1/"
where:
x = weight of solute adsorbed
m = weight of carbon
C = equilibrium concentration of solute in solution after adsorption
K and n are constants
The isotherms are normally plotted on a log-log scale. The extrapolation of the isotherm line
to the initial concentration (abscissa) gives the theoretical adsorption capacity of that
carbon when it is in equilibrium with the influent concentration.
The advantage of isotherms are: (1) they are relatively simple tests to perform; (2) they
indicate whether the desired degree of treatment can be readily achieved; and (3) they give
the approximate adsorptive capacity of the carbon in a column application. However,
isotherm results should not be used to extrapolate carbon capacities and dosages to full-scale
plant size.
Typical isotherms obtained for the same carbon with different secondary wastewater
effluents were reported by Masse (37) (see Figure 7-5). The results shown on Figure 7-5
indicate that the adsorptive capacity of the carbon with respect to COD varies from 0.37 to
0.12 Ib of COD/lb of carbon. This is equivalent to 1.58 to 1.09 Ib of carbon/1,000 gal of
throughput. The carbon requirements (per unit volume of wastewater treated) obtained
from isotherms are conservative (i.e., high), because removal by adsorption alone is
estimated.
7-30
-------
FIGURE 7-5
71
co
CJ
BO
XIS
0.0
COD ISOTHERMS USING VIRGIN CARBON
AND DIFFERENT SECONDARY WASTEWATER EFFLUENTS (37)
(j)C0=0.37
(C) RESIDUAL COD CONC., MG/L
-------
Since isotherms cannot measure the quantity of organics removed by filtration and
biological action, pilot column testing must be conducted to evaluate the effect of these
factors. Column testing helps to determine: (1) the required contact time; (2) the
adsorptive capacity of the carbon; (3) the pressure drop across the beds and backwash
requirements for downflow operation; and (4) the shape of the column exhaustion wave
front.
The column used in pilot testing should have a diameter of at least 4 inches. The depth of
column depends on the range of contact times being considered, as does the hydraulic
loading. Normally, two to four columns are used in series, since this arrangement permits
evaluation of the effect of different contact times on effluent quality. When loading
granular carbon into the test column (a "wet" packing procedure is recommended), care
must be exercised to avoid entrapping air within the carbon column. Air entrapment causes
channeling and reduces the contact area, which in turn yields false test results. The Process
Design Manual for Carbon Adsorption (2) describes the conduct of pilot operations in more
detail.
7.5.3 Costs of Tertiary Carbon Treatment
The capital cost of tertiary granular carbon systems will vary widely depending on the
particular system design and pretreatment provided. Direct application of secondary
effluent to downflow carbon adsorption columns as practiced at Pomona, California, will
result in a smaller capital investment than a tertiary system using effluent polishing prior to
carbon treatment. Total capital costs for carbon treatment may be estimated from
Figure 7-6 for two levels of carbon dosage. For tertiary treatment, the carbon dosage will
generally be in the 200 to 500 Ib/mil gal range.
Operating cost will depend primarily on the organic loading and associated carbon dosage.
Annual costs for power, labor, and maintenance may be estimated from Figure 7-7 which is
based on the experience at Lake Tahoe. Total annual and unit costs may be estimated from
Figure 7-8 for various carbon dosages.
7-32
-------
FIGURE 7-6
TOTAL CAPITAL COSTS FOR CARBON TREATMENT (2)
1.000,
9
8
7;
6
100
9
8
7
6
5
10
9
8
7
6
5
NOTES
1. EPA STP INDEX = 175
2. COSTS INCLUDE 100 PERCENT
EXCESS REGENERATION
CAPACITY ABOVE AVERAGE
ESTIMATED DOSAGE.
3. COSTS INCLUDE PUMP STATION,
CONTACTOR SYSTEM, REGENERA-
TION SYSTEM, BUILDINGS,
ENGINEERING, LEGAL, INTEREST
CHARGES.
1200
200
CARBON
DOSAGE .
LB/MIL GAL
4 5 6 7 8 9 10 2 3
DESIGN AVERAGE FLOW, MGD
4 5678 9100
7-33
-------
1.000
9
8
7
6
FIGURE 7-7
CARBON ADSORPTION
OPERATION AND MAINTENANCE COSTS (2)
100
9
8
7
10
9
8
7
6
YEAR - 1972
LABOR PAYROLL COST - $5/HR
POWER COST - $.02/kWh
456789 10 2 3
AVERAGE DESIGN FLOW, MGD
5 6 7 8 9 100
7-34
-------
FIGURE 7-8
TOTAL ANNUAL AND UNIT COSTS FOR CARBON TREATMENT (2)
10,0
1,000
9
8
7
100
9
8
7
6
10
NOTES
1, COSTS INCLUDE OPERATION, 4.
MAINTENANCE, AND CAPITAL
AMORTIZATION AT 51/2*
FOR 25 YEARS.
2. EPA STP INDEX = 175.0 R
U
3. COSTS INCLUDE 100 PERCENT
EXCESS REGENERATION
CAPACITY RELATIVE TO
ESTIMATED AVERAGE
DOSAGE.
CAPITAL COSTS INCLUDE
PUMP STATIONS, CONTACTORS,
REGENERATION SYSTEM,
BUILDINGS, ENGINEERING,
LEGAL. INTEREST CHARGES.
O&M COSTS INCLUDE LABOR,
POWER, FUEL, MAKEUP
CARBON, MAINTENANCE
MATERIALS.
1200
. CARBON
800t DOSAGE
400 ( LBS/MIL GAL
200
1201U
800 / CARBON
400 / DOSAGE
2QOJ LBS/MIL GAL
200
80
70
60
50
40
30
20
10
3 45678910 2 3 4 56789 100
DESIGN AVERAGE FLOW. MGD
7-35
-------
7.6 References
1. Process Design Manual for Suspended Sotids Removal. Office of Technology Transfer,
U. S. Environmental Protection Agency, Washington, D. C. (1974).
2. Process Design Manual for Carbon Adsorption. Office of Technology Transfer,
U. S. Environmental Protection Agency, Washington, D. C. (October, 1973).
3. Metcalf & Eddy Engineers, Report to Marlboro, Massachusetts, on "Operation of
Sewage Treatment Works for the Years 1961 and 1962." (August 19, 1963).
4. Loehr, R., and Stephenson, R., An Oxidation Pond as a Tertiary Treatment Device.
Journal of the Sanitary Engineering Division, ASCE 91, No. 3, pp. 31-44 (1965).
5. Private Communication with James Neighbor, Vice President, Hinde Engineering
Company, Highland Park, Illinois (October 28, 1970).
6. Eckenfelder, W. W., Engineering Aspects of Surface Aerator Design. Presented at the
22nd Industrial Waste Conference, Purdue University (May, 1967).
7. Edde, G., Field Research Studies of Hydraulic Mixing Patterns in Mechanically Aerated
Stabilization Basins. Presented at the International Congress in Industrial Wastewater,
Stockholm, Sweden (November, 1970).
8. Fall, E., Retention Pond Improves Activated Sludge Effluent Quality. Journal Water
Pollution Control Federation, 37, No. 9, pp. 1,194-1,202 (1965).
9. Reynolds, Jeremiah, Decatur Tertiary Treatment Plan Proves its Worth. Water and
Sewage Works, 115, No. 12, pp. 553-584 (1968).
10. Hickman, Paul, Polishing and Secondary Effluents and Treatment Bypasses. Presented
at the 26th Industrial Waste Conference, Purdue University (May 4, 1971).
11. Bodien, D. G., and Stenburg, R. L., Microstraining Effectively Polishes Activated
Sludge Plant Effluent. Water and Wastes Engineering, 3, No. 9, pp. 74-77 (1966).
12. Diaper, E.W.J., Tertiary Treatment by Microstraining. Water and Sewage Works, 115,
No. 6, pp. 202-207 (1969).
13. Kreissl, J. F., Granular Media Filtration of Wastewater: An Assessment. Presented at
Seminar "Filtration of Water and Wastewater," Ann Arbor, Michigan (January, 1973).
7-36
-------
14. Hirsch, A. A., Backwash Investigation of a Proposed Simple Uniformity Control.
Journal AWWA, 60, 570 (May, 1968).
15. Neptune Microfloc Technical Release No. 4, Design Helps for Tertiary Filtration.
16. University of Michigan short course, January 25-26, 1973. Reported by Thomas
Hoogerhyde, Michigan Department of Health.
17. Private communication with H. M. Mueller, Jr., Neptune Microfloc (April, 1973).
18. Private communication with S. T. Welch, Superintendent Spring Creek Pollution
Control Facility (April, 1973).
19. Private communication with J. Wiley Finney, Jr., Treatment results Hite Creek Tertiary
Plant, Louisville, Kentucky (April, 1973).
20. Gulp, G. L., and Hansen, S. P., Extended Aeration Effluent Polishing by Mixed-Media
Filtration. Water and Sewage Works, 114, No. 2, pp. 46-51 (1967).
21. Technical Bulletin, Dravo Corp., Water and Waste Treatment Division, Pittsburgh, Pa.,
Ventura, California East Side STP Test Report.
22. Zenz, D. R., Weingarden, M. J., and Bogusch, E.D., Hanover Park Experimental Bay
Project (March 8, 1972).
23. Convery, J. J., Solids Removal Processes. Nutrient Removal and Advanced Waste
Treatment Symposium. Presented by Federal Water Pollution Control Administration,
Cincinnati, Ohio (April 29-30, 1969).
24. Laverty, F. B., Stone, R. and Meyerson, L. A., Reclaiming Hyperion Effluent. Journal
Sanitary Engineering Division, ASCE, 87, 6, 1 (November, 1961).
25. Lynam, B., Ettelt, G., and McAloon, T. J., Tertiary Treatment at Metro Chicago by
Means of Rapid Sand Filtration and Microstrainers. Journal Water Pollution Control
Federation, 41, p. 247 (February, 1969).
26. Performance Data Contained in Hydroclear Corporation Catalogue, Avon Lake, Ohio,
as tested by the Clark County Utilities Department, Springfield, Ohio (May, 1969).
27. U. S. EPA Internal Monthly Reports, Piscataway, Md. (March-September, 1973).
28. U. S. EPA Internal Monthly Reports, Ely, Minn. (April-December, 1973).
7-37
-------
29. Study of Upflow Filter for Tertiary Treatment. U. S. EPA, Project No. 17030 DMA
(August, 1972).
30. Berg, E. L., and Brunner, C. A., Pressure Filtration of Secondary Treatment Plant
Effluent. Water and Wastes Engineering, p. 54 (October, 1969).
31. U. S. EPA Internal Monthly Report, Pomona Pilot Plant, summarizing previous work
(June, 1972).
32. Appraisal of Granular Carbon Contacting, Report No. TWRC 11 and 12, Federal Water
Pollution Control Administration, Ohio Basin Region, Cincinnati, Ohio (May, 1969).
33. Parkhurst, J. D., Dryden, F. D., McDermott, G. N. and English, J., Pomona Activated
Carbon Pilot Plant, Journal Water Pollution Control Federation, 39, No. 10, Part 2, pp.
R70-R81 (1967).
34. English, J., Masse, H. N., Carry, C. W., Pitkin, J. B., and Haskins, J. E., Removal of
Organics from Wastewater by Activated Carbon. Water, 67, No. 107, pp 147-153
(1970).
35. Gulp, R. L., and Gulp, G. L., Advanced Wastewater Treatment. New York: Van
Nostrand-Reinhold Company (1971).
36. Weber, W. J., Jr., Hopkins, C. B., and Bloom, R., Jr., Expanded-Bed Active-Carbon
Adsorption Systems for Wastewater Treatment. In Water Quality Improvement by
Physical and Chemical Processes. Edited by Gloyna, E. F., and Eckenfelder, W. W., Jr.,
University of Texas Press, Austin, pp. 294-311 (1970).
37. Masse, A. N., Organic Residue Removal. Nutrient Removal and Advanced Waste
Treatment Symposium. Presented by Federal Water Pollution Control Administration,
Cincinnati, Ohio (April 29-30, 1969)
7-38
-------
CHAPTER 8
PREAERATION AND POSTAERATION PRACTICES
8.1 Preaeration
Preaeration of wastewater has been practiced for over 50 years throughout the United
States, generally for the purpose of odor control and to improve the treatability of the
wastewater. Short aeration periods ranging up to 15 minutes have been found adequate for
these purposes. For longer aeration periods, the additional benefits of grease separation and
improved flocculation of solids have also been observed (1).
While the use of aerated grit chambers is becoming increasingly popular as a pretreatment
unit in wastewater treatment plants, their use should not be expected to substantially
increase BOD or SS removal during primary clarification due to the relatively short
detention times normally employed.
8.1.1 Preaeration Process Design Considerations
The major parameters to be considered in the design of preaeration facilities are rate of air
application and detention time. In order to maintain proper agitation, the air supply system
should provide a range of 1.0 to 4.0 cfm per lineal foot of tank. This range will assure
adequate performance for nearly all physical tank layouts and types of aeration equipment
used.
Effective preaeration has been achieved at detention times of 45 minutes and less (1) (2).
The Ten-States Standards (3) recommends a detention time of 30 minutes for effective
solids flocculation when inorganic chemicals are used in conjunction with preaeration. For
appreciable BOD reduction, a minimum of 45 minutes is recommended. In addition, the use
of polyelectrolytes may vary these detention times.
8.1.2 Upgrading Considerations for Preaeration
Several efforts have been made to determine the effects of preaeration on primary clarifier
performance. In 1961, Seidel and Baumann (2) conducted a study at the Ames, Iowa,
secondary treatment plant to determine the effects of preaeration by direct comparison
with a parallel primary clarifier receiving an equal flow of the same waste without
preaeration. They determined that with 45 minutes detention and an aeration rate of
0.1 cu ft/gal of wastewater, BOD and SS removals were both increased by seven to eight
percent in the primary tank. In this instance, an economic evaluation based on conventional
design standards showed that preaeration costs increase the total annual plant operating cost
by two to three percent.
8-1
-------
In an analysis of operation data from 38 plants using preaeration, Roe (1) observed
increased primary clarifier SS removals were obtained in some instances. Quantities of air
used varied between 0.06 and 0.15 cu ft/gal. However, informal tests at the Allegheny
County Sanitary Authority Wastewater Treatment Plant showed that little or no
improvement in primary clarifier performance was achieved using preaeration with a
detention time of 45 minutes.
These results and others as reported by Eliassen (4) have lead to some disagreement as to the
desirability of providing preaeration for increased primary SS or BOD removals. Although
the studies by Roe, and by Seidel and Baumann have established that primary clarifier
performance will improve with prolonged preaeration time, it is often difficult to justify the
additional cost since the incremental removals achieved as a result of preaeration probably
would be partially or completely obtained in the secondary system without preaeration.
The specific conditions involved in determining the merits of preaeration must be evaluated
for each treatment situation. The costs of providing sufficient preaeration and clarifier
detention times required for significant improvements in primary performance must be
weighed against the savings that would be derived from the resulting reduction in secondary
loading.
The most likely upgrading situations involving preaeration that will be encountered are: (1)
when an existing plant with preaeration is to be upgraded, and (2) when considering
chemical additions prior to existing primary treatment. In the latter case, existing plant
hydraulics will have a significant impact on the designer's options.
Where existing preaeration-primary plants are being upgraded to secondary treatment, it
may be beneficial to inject waste-activated sludge upstream of preaeration to improve
primary clarifier performance.
8.1.3 Preaeration with Chemical Addition
Removal of BOD, SS and phosphorus in primary clarification may be substantially improved
through the use of chemical additions to preaeration units (5). The addition of chemicals
such as lime, alum, ferric chloride, or organic polyelectrolytes to the wastewater flow ahead
of the preaeration unit allows for adequate mixing of the chemical by air stirring. Chemical
addition can also reduce the preaeration time required to obtain improved primary clarifier
performance. Before implementing preaeration with chemical addition, it is recommended
that full- or pilot-scale tests be conducted to determine optimum chemical dosages and
removal efficiencies.
The Central Contra Costa Sanitary District near San Francisco, California, has operated a
2-mgd advanced treatment test facility since November 1971. The existing preaeration and
primary clarification facilities shown on Figure 8-1 were upgraded by providing lime
8-2
-------
FIGURE 8-1
CD
CO
PRIMARY TREATMENT UNITS AT THE
CENTRAL CONTRA COSTA SANITARY DISTRICT WATER RECLAMATION PLANT
AGITATION AIR
HEADER
HELICAL SCUM SKIMMER
PREAERATION,
FLOCCULATION AND
GRIT REMOVAL
TANK
LONGITUDINAL
SLUDGE COLLECTOR
LUDGE RECIRCULATION
LINE
NFLUENT
SLUDGE CROSS COLLECTOR
-------
addition to precipitate phosphorus and to maximize the removal of raw wastewater solids
thus reducing the load on the subsequent combined oxidation-nitrification unit (5) (6).
At the Contra Costa facility, lime is added to the raw wastewater in a steep channel
immediately after screening. A hydraulic jump occurs ahead of the preaeration tank and
provides the necessary agitation for chemical mixing. Parker and Niles (5) indicate that the
combined preaeration and grit removal tank performs well as a flocculation basin using lime
as the coagulant. Diffused air mixing is used to promote the formation of large readily
settleable flocculent particles. To further enhance coagulation and floe aggregation, settled
primary solids are recirculated to the preaeration tank influent. Increasing the solids
concentration and solids contact in this manner has been shown to improve flocculation
efficiency.
Air diffusers of the swing-type were selected for this application as being most compatible
with the multiple function served by the preaeration tank. Diffused air induced currents
provide a readily adjustable stirring action necessary to promote flocculation and grit
collection. Problems with scaling have often been encountered where lime addition is
practiced and rag fouling of mechanical pretreatment equipment is common. The easy
accessibility of air diffusers for maintenance minimizes these problems.
Critical design criteria for flocculation are detention time and aeration rate. At Contra
Costa, the preaeration tank was designed for 30 minutes detention time at average
dry-weather flow. Satisfactory results have been obtained at operational detention times of
20 minutes. Air supply rates for lime flocculation must be lower than the rates used for
conventional preaeration to avoid shearing of the floe. Although optimal rates have not been
definitely established, preliminary results based on the Contra Costa experience indicate air
requirements may be 1/3 to 1/7 those normally used for preaeration.
At an overflow rate of 1,300 gpd/sq ft and 1.5 hours detention, the preaerated, chemically
treated system significantly outperforms a parallel conventional system, as shown in
Table 8-1 (6).
8.2 Postaeration
Many states are considering or have already enacted legislation requiring the maintenance of
minimum DO concentrations in wastewater treatment plant effluents. This design criteria is
established to provide additional oxygen to the receiving waters. Plant effluents from
secondary clarifiers normally contain between 0.5 to 2.0 mg/1 of DO. Most surface water
quality standards, depending on the intended water use, specify a minimum DO
concentration of 4.0 mg/1.
8-4
-------
TABLE 8-1
PERFORMANCE OF PRIMARY FACILITIES
AT CENTRAL CONTRA COSTA SANITARY DISTRICT
WITH CHEMICAL TREATMENT TO PREAERATION
pH 11.5 Operation pH 11.0 Operation
Ca(OH)2 = 500 mg/1 Ca(QH)2 = 400 mg/1
Control Chemical Control Chemical
Primary Primary Primary Primary
BOD Removal, Percent 46 74 37 69
SS Removal, Percent 71 79 71 76
Grease Removal, Percent 44 79 21 64
8.2.1 Postaeration Process and Design Considerations
There are at least four methods available for the postaeration of a wastewater treatment
plant effluent. These are shown on Figure 8-2. Most of these devices were initially developed
for water treatment and are now being used in the wastewater treatment field.
8.2.1.1 Diffused Aeration
Diffused air aerators are usually placed in concrete tanks which are commonly 9 to 15 feet
deep and 10 to 30 feet wide. Ratios of width to depth of 1.5 are most desirable and should
not exceed 2.0 if effective mixing is to be obtained. Tank length is governed by the desired
detention period, which usually varies from 10 to 30 minutes.
Aeration systems are designed on the basis of their oxygen-transfer rate at standard
conditions. Standard conditions are defined as 1.0 atmosphere dry pressure at 20 deg C for
tap water containing 0.0 mg/1 DO. The required rate at which oxygen must be transferred to
the wastewater to raise the DO the desired amount under actual operating conditions is
given by the following:
AOR = 8.33Q(C-C0)
where:
AOR = Actual oxygen-transfer rate, Ib O2/day
Q = Postaeration influent flow, mgd
C = Required final DO level after postaeration, mg/1
Co = DO concentration of the postaeration influent, mg/1
8-5
-------
FIGURE 8-2
VARIOUS POSTAERATION DEVICES
A. DIFFUSED AERATION
OXYGEN SOURCE OR
MR COMPRESSOR^
o
B. MECHANICAL AERATION
UULTU
B-1 TURBINE TYPE AERATOR
C. CASCADE A-ERATION
HEAD
r^Z^LOSS
D. U-TUBE AERATION
AIR,. , VENTURI
ASPIRATOR
EFFLUENT
CHANNEL
B-2 PUMP TYPE AERATOR
BINE-^
AIR LINE
._ /SPARGER
p/. )
K
B-3 AGITATOR SPARGED-SYSTEM
8-6
-------
This actual oxygen-transfer rate may be adjusted to standard conditions by applying
correction factors according to the following equation (7):
SOR =
AOR
3CS-C\
k C20/
1.024T-20 a
where:
SOR = Standard oxygen-transfer rate, Ib 02/day
Cs - DO saturation concentration of tap water at
temperature T, mg/1
20 = DO saturation concentration of tap water at
20 deg C, mg/1
T = Design temperature of the wastewater, deg C
a = 02 transfer coefficient of the wastewater
3 = 02 saturation coefficient of the wastewater
Diffused air systems are designed to provide firm blower capacity, which is the capacity
remaining with the largest blower out of service. The maximum air rate required to provide
firm capacity may be computed from the standard oxygen-transfer rate as follows:
A = SOR
1440 et Ya P0
where:
Am = Firm blower capacity, cfm
SOR = Standard oxygen-transfer rate, Ib 02/day
et ~ ^2 transfer efficiency of 02 in diffused air
^a = Specific weight of air at design temperature and
relative humidity, pcf
Po = 02 content of dry air, proportion by weight
The use of oxygen aeration in the activated sludge process may eliminate the need for
postaeration. Oxygen-aerated mixed liquor discharged to the secondary clarifier usually has
a DO of at least 6.0 mg/1 (8). In some instances, postaeration with high purity oxygen may
be advantageous.
The Reno-Sparks, Nevada Joint Wastewater Treatment Facility provides activated sludge
secondary treatment for an average design flow of 20 mgd. State standards requiring that a
minimum plant effluent DO concentration of 6.0 mg/1 be maintained are met by diffused air
postaeration of the secondary clarifier effluent prior to chlorine disinfection and discharge
8-7
-------
over a final effluent weir (9). The postaeration unit consists of a single 210-foot by 30-foot
by 15-foot tank that provides a detention time of 48 minutes at average design flow. The
diffused air system serves both secondary and postaeration requirements using two blowers
with a total capacity range of 7,240 to 16,080 cfm of standard air. The system is regulated
on the basis of the secondary aeration air requirements with surplus air used for
postaeration.
Examination of plant operating records for the year 1972 showed that an average of
17.0 mgd was treated. Monthly average secondary clarifier effluent DO concentration
ranged from 0.3 mg/1 to 0.6 mg/1, while the plant discharge DO concentration was observed
to average between 6.1 mg/1 and 9.4 mg/1.
A recent study was conducted to determine on a daily basis the increase in the DO
concentration of the treated effluent due to the postaeration facility alone. Typical data
taken during this study are as follows:
Flow Rate, Q = 18.21 mgd
Wastewater temperature, T = 20° C
Postaeration influent DO, Co = 0.5 mg/1
Air supply rate, Am = 6,050 cfm
Observed postaeration DO, C = 7.6 mg/1
Assuming a diffuser oxygen-transfer efficiency, e^, of 10 percent; and oxygen transfer and
oxygen saturation coefficients, a and 3 of 0.85 and 0.95, respectively; the expected
postaeration effluent DO concentration, C, may be computed using the equations stated
previously. The computed value for this data is 7.4 mg/1, which is in close agreement with
the observed DO concentration of 7.6 mg/1.
8.2.1.2 Mechanical Aeration
Mechanical aerators are generally grouped in two broad categories: turbine types and pump
types, as shown on Figure 8-2. In all types, oxygen transfer occurs through a vortexing
action and/or from the interfacial exposure of large volumes of liquid sprayed over the
surface.
To optimize aeration and mixing and to avoid interference between units, aerator
manufacturers have developed criteria for minimum areas and depths, depending on the
horsepower of the aerator and the configuration of the impeller.
Mechanical aeration systems are designed on the basis of the horsepower required to
produce the needed standard oxygen transfer rate (SOR), which was defined in the previous
section. One aeration design equation proposed by Kormanik for a postaeration basin is
(10):
8-8
-------
p= SOR
24 NO ₯ n
&
where:
P = Horsepower required
SOR = Standard oxygen-transfer rate, Ib 02/day
N0 =02 transfer efficiency under standard conditions in tap
water, Ib O2/hp-hr
₯ = Correction factor related to basin geometry
Tl = Aerator efficiency correction
8.2.1.3 Cascade Aeration
Cascade aeration takes advantage of the effluent discharge to create a series of steps or weirs
over which the flow moves in fairly thin layers, as shown on Figure 8-2. The objective is the
maximization of turbulence to increase oxygen transfer. Head requirements vary from three
to ten feet, depending upon the initial DO and the desired increase. If the necessary head is
not available, effluent pumping is required.
In England, the Water Research Laboratory has performed investigations to qualify as much
as possible the layout of cascade aeration schemes. Barrett and others proposed the
following formulae (11):
r = (Cs - C0)/(C8 - C)
h = r-1
0.11ab(l-t-0.046T)
where:
r - The deficit ratio
Cs - DO saturation concentration of the wastewater at
temperature T, mg/1
C0 = DO concentration of the postaeration influent, mg/1
C = Required final DO level after postaeration, mg/1
a = Water quality parameter equal to 0.8 for a wastewater
treatment plant effluent
b = Weir geometry parameter equal to unity for a free
weir and 1.3 for the step weirs used in their
experimental work
T = Water temperature in deg C
h = Height in feet through which water falls
8-9
-------
For example, to raise the DO concentration of a wastewater treatment plant effluent from
0.5 mg/1 to 4.5 mg/1 at 20 deg C, the computed overall height requirement for a series of
step weirs would be approximately 4.0 feet. However, it should be pointed out that the
values of the parameters a and b are somewhat arbitrary and need further refinement to
substantiate preliminary results.
The wastewater treatment plant at Pittsfield, Massachusetts, a trickling filter plant, is
followed by two cascades in parallel, each receiving half of the effluent flow. Each cascade
consists of a series of 18 three-foot wide concrete steps with a total horizontal length of
24.6 feet and a total vertical drop of 10.3 feet, as shown on Figure 8-3.
Recently, a study was conducted to determine the increase in the effluent DO concentration
resulting from flow over the cascades (12). During the study period the total treated flow
averaged 7.0 mgd at ISdegC. The secondary clarifier effluent DO concentration was
observed to vary from 3.9 to 4.2 mg/1, while the aerated final effluent DO varied between
5.8 and 6.2 mg/1. The values computed using Barrett's formula with a = 0.8 and b = 1.1 were
six to ten percent higher than the observed final effluent DO concentrations for the
corresponding clarifier effluent and temperature. This value of b is reasonable in this
instance since the concrete steps involved would be expected to produce more turbulence
than a free weir (b = 1.0), but less turbulence than step weirs (b = 1.3).
8.2.1.4 U-Tube Aeration
The U-tube aerator consists of two basic components: a conduit to provide a vertical
U-shaped flow path and a device for entraining air into the stream flow in the down leg of
the conduit as indicated on Figure 8-2. The entrainment device is one of two types: (1)
aspirator; or (2) compressor and diffuser. In either case, the entrained air is carried along the
down leg of the tube because the water velocity exceeds the buoyant rising velocity of the
air bubbles.
Various design considerations include air-to-water ratio, tube cross-sectional area, and depth.
The maximum air-to-water ratio practicable is a function of the velocity through the system.
At velocities of approximately 4 fps, 20 percent air-to-water injection is about the limit for
satisfactory operation (13). The hydraulic head requirements for plants of 5 mgd or less
should be less than five feet. If sufficient head is not available, the flow may be pumped
through the U-tube.
Speece and Orosco (13) have suggested that one economic method of construction for deep
U-tubes, greater than 20 feet in depth, would be a circular hole bored into the soil. The hole
would be cased and a smaller pipe then suspended a few feet from the bottom of the hole as
shown on Figure 8-2. The diameter of the smaller pipe is selected so that its cross-sectional
area is approximately equal to the cross-sectional area of the annular space between the two
pipes. Thus, the velocity of the water will be approximately equal in both legs of the
U-tube.
8-10
-------
CO
FIGURE 8-3
CASCADE AERATOR AT PITTSFIELD,MASS.
12 STEPS
6" DROP
18" TREAD
6 STEPS
85/8" DROP
IB" TREAD
-------
Presently there are no known U-tube installations in wastewater treatment plants for
postaeration. However, full-scale (8- to 20-inch diameter) U-tube units are now being
operated to provide in situ aeration in sanitary force mains in two communities in Louisiana
and Texas (14). These installations have been effective in reducing serious odor and
corrosion problems resulting from sulfides. No maintenance has been required in
approximately two years of continuous operation. The possibilities of using a U-tube as a
postaeration device seem good at this time.
8.2.2 Upgrading Considerations for Postaeration
The selection of postaeration facilities for treatment plant upgrading is greatly influenced by
the specific conditions encountered. The degree to which existing facilities can be used in
the upgrading will, in many cases, determine the proper selection of equipment. For
example, where the plant is undergoing general upgrading, it may be possible to
substantially reduce the costs of mechanical or diffused air facilities by converting an
existing basin for use as a postaeration tank.
In some situations there may be operational and/or other advantages in the use of the same
type of aeration system for postaeration as is already in use elsewhere in the plant. This is
especially true of a diffused air secondary treatment plant where the same blowers and
housing structure could be used to supply the postaeration unit as well, with only minor
modification.
Cascade postaeration has the obvious advantage of no power, labor or maintenance costs
where sufficient head is available between the plant discharge and the receiving water. Where
stream standards rather than effluent standards are applicable, but available head is limited
at high river levels, it may be possible to use cascade postaeration by taking advantage of
seasonal fluctuations in the dilution capacity of the stream. The plant effluent DO is most
critical to stream quality during low flow periods. This coincides with the time when the
maximum head is available for effective cascade postaeration of the effluent. At higher
stream flows when the cascade may be partially submerged, less postaeration is required
because of greater dilution.
Although U-tube aeration has been utilized only for sanitary force main aeration to date,
the use of this technique for postaeration is promising where sufficient head is available for
gravity flow. U-tube postaeration appears particularly attractive where limited space is
available for facilities.
Because the peculiarities of each upgrading situation so greatly affect the feasibility and
costs of the various alternative postaeration devices, no clear picture emerges as to the
relative merits and costs of each method. In each upgrading, an analysis should be made of
the applicable conditions such as type of treatment involved, usable existing facilities, and
8-12
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available tank volume and head. As will be discussed in Chapter 9, it may be feasible to
combine postaeration and chlorine disinfection advantageously.
It must be noted that postaeration can be expected to produce a detergent type foam. Thus,
the facilities should be designed to retain and destroy the foam created, rather than to allow
it to flow unimpeded to the surface of the receiving water.
8.3 References
1. Roe, F., Pre-aeration and Air Flocculation. Journal Water Pollution Control
Federation, 23, No. 2, pp. 127-140 (1951).
2. Seidel, H., and Baumann, E., Effect of Pre-aeration on the Primary Treatment of
Sewage. Journal Water Pollution Control Federation, 33, No. 4, pp. 339-355 (1961).
3. Recommended Standards for Sewage Works. Great Lakes - Upper Mississippi River
Board of State Sanitary Engineers (1971).
4. Eliassen, R. and Coburn, D.F., Versatility and Expandability ofPretreatment, Journal
of the Sanitary Engineering Division, ASCE, 95, No. 2, pp. 299-310 (1969).
5. Parker, D.S. and Niles, D.G., Full-Scale Test Plant at Contra Costa Turns Out Valuable
Data on Advanced Treatment. Bulletin of the California Water Pollution Control
Association, 9, No. 1 (1972).
6. Horstkotte, G.A., Niles, D.G., Parker, D.S., and Caldwell, D.H., Full-Scale Testing of a
Water Reclamation System. Journal Water Pollution Control Federation, 46 No. 1,
pp. 181-197(1974).
7. Metcalf & Eddy, Inc., Wastewater Engineering: Collection, Treatment, Disposal.
McGraw-Hill, Inc. (1972).
8. Albertsson, J., et al, Investigation of the Use of High Purity Oxygen Aeration in the
Conventional Activated Sludge Process. Federal Water Quality Administration,
Program Number 17050 DNW, (1970).
9. Private communication with G. Davis, Superintendent, Reno-Sparks Joint Treatment
Plant, Sparks, Nevada, November, 1973.
10. Kormanik, R., Simplified Mathematical Procedure for Designing Post Aeration
Systems. Journal Water Pollution Control Federation, 41, No. 11, pp. 1956-1958
(1969).
8-13
-------
11. Barrett, M.J., et al, Aeration Studies of Four Weir Systems. Water and Water
Engineering, 64, No. 9, pp. 407-413 (1960).
12. Private communication with W. Fallen, Plant Superintendent, Pittsfield, Massachusetts,
October, 1973.
13. Speece, R., and Orosco, R., Design of U-tube Aeration Systems. Journal of the
Sanitary Engineering Division, ASCE, 96, No. 3, pp. 715-726 (1970).
14. Mitchell, R.C., U-Tube Aeration. U. S. EPA, Project No. 17050 DVT, Contract No.
68-01-0120 (September, 1973).
8-14
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CHAPTER 9
DISINFECTION AND ODOR CONTROL
9.1 General
Disinfection and odor control are receiving increased attention from regulatory agencies
through the establishment and enforcement of rigid bacteriological effluent standards and
air pollution standards. In upgrading situations, the need for improved disinfection and the
elimination of odor problems are frequently encountered. Adequate and reliable
disinfection and odor control are essential to ensure that wastewater treatment plants are
environmentally safe and aesthetically acceptable to the public.
9.2 Disinfection
In 1973, 48 percent of all municipal wastewater treatment plants in the United States were
equipped with chlorination facilities for disinfection purposes (1). Other disinfectants, such
as ozone, ultraviolet light and bromine chloride, are currently being evaluated by the U. S.
EPA to determine their potential use in disinfecting wastewater treatment plant effluents.
Recent shortages and price escalation of liquid chlorine have emphasized the need to
consider other methods of disinfection.
9.2.1 Chlorine Fundamentals
When chlorine is added to water, it hydrolyzes to form hypochlorous acid (HOC1) as
follows:
ci2 + H2o ^ HOCI + H+ + cr
The hypochlorous acid further ionizes to form hypochlorite ion (OC1") as follows:
The equilibrium concentrations of HOCI and OC1" in solution depend on the pH of the
wastewater. Increasing the pH will shift the equilibrium toward the formation of higher
concentrations of OC1" and, conversely, decreasing the pH will shift the equilibrium toward
the formation of higher concentrations of HOCI.
Chlorine may also be applied as sodium hypochlorite (NaOCl), or calcium hypochlorite
[Ca (OC1)2]. The hypochlorite form reacts as follows:
NaOCl - s- Na+ + OC1"
9-1
-------
Ca (OC1)2 - * Ca2+ 2OC1'
H+ + OCr^^ HOC1
The quantity of HOC1 plus OC1" in wastewater is known as free available chlorine.
Hypochlorous acid is an extremely potent germicide ai pH 6.5 to 7.5 (2). It is also a very
active oxidizing agent and is therefore short lived in the presence of readily oxidized
compounds such as ammonia. Since most wastewater effluents contain ammonia, the
following reactions will occur upon the addition of chlorine:
NH3 + HOC1 - *- NH2 Cl + H20
(monochloramine)
NH2 Cl + HOC1 - >NHC12 + H2O
(dichloramine)
NHC12 + HOC1 - *- NC13 + H20
(nitrogentrichloride)
The two species that predominate in most cases are monochloramine and dichloramine.
They are commonly referred to as the combined available chlorine. The reaction with
ammonia is unfortunate, because chloramines are many times less potent than hypochlorous
acid as a disinfectant. Figure 9-1 illustrates the relative potency of free and combined forms
of chlorine.
Hypochlorous acid also reacts with other organic matter in wastewater such as amino acids
and inorganic matter such as sulfites and nitrites to produce chlorine compounds that have
very little or no disinfecting power (2). Design engineers should be aware of the extent of
such side reactions when determining the optimum chlorine dosages to apply to a
wastewater.
Total chlorine residual is the sum of the combined and free chlorine concentrations
remaining after a specified period of time. Chlorine demand is defined as the difference
between the chlorine applied and the total residual chlorine remaining at the end of the
contact period (3). It provides a measure of all chlorine demanding reactions including
disinfection.
Browning and McLaren (3) analyzed 844 samples from 12 primary treatment plants and 777
samples from 15 secondary treatment plants to obtain a correlation between residual
chlorine and coliform MPN per 100ml. The results of their analyses are shown on
Figure 9-2. The curves represent the MPN remaining after 30 minutes of chlorine contact in
9-2
-------
FIGURE 9-1
RELATIONSHIP BETWEEN CONCENTRATION AND TIME
FOR 99 PERCENT DESTRUCTION OF ESCHERICHIA CPU
BY DIFFERENT FORMS OF CHLORINE AT 2 TO 6°C (2)
10
1 .0-
0.10-
co
oe
0.010 -
I _ III I I 1-1
I I I-III
\ X
\\
\ \
\
HYPOCHLOROUS
ACID (HOCI)
-HYPOCHLORITE ION
(OCD
MONOCHLORAMINE
(NH2CI)
\
0. 001
5 10
50 100
500 1000
MINUTES
9-3
-------
FIGURE 9-2
MPN COLIFORM VS. CHLORINE RESIDUAL (3)
100,000-3
0.000-
1000_
00-
10
I I I I I
CONTACT TIME OF 30 MINUTES
o - PRIMARY EFFLUENT
v - SECONDARY EFFLUENT
T
0123456
CHLORINE RESIDUAL, MG/L MODIFIED STARCH - IODIDE METHOD
9-4
-------
a well-designed chlorine contact chamber. Insufficient data were available for residuals
above 4 mg/1. The results of these analyses should not be quantitatively applied to any
primary or secondary effluent. However, the study provides a good indication of the
effectiveness of increased chlorine residuals on primary and secondary effluents.
Other factors which influence the germicidal effectiveness of chlorine are pH and
wastewater temperature. At pH values greater than 7.5, the less potent form of free chlorine
(OC1~) predominates. Increased pH also diminishes the disinfecting efficiency of
monochloramine (2). It has also been demonstrated that the germicidal effectiveness of free
and combined chlorine is markedly diminished with decreasing wastewater temperature (2).
9.2.2 Chlorine Dosage Rates
The following table was taken from the Water Pollution Control Federation's Sewage
Treatment Plant Design Manual of Practice No. 8 and contains ranges of chlorine dosages
recommended for disinfection (4).
TABLE 9-1
CHLORINE DOSAGE RANGES
Waste Chlorine Dosage
mg/1
Raw Sewage 6 to 12
Raw Sewage (Septic) 12 to 25
Settled Sewage 5 to 10
Settled Sewage (Septic) 12 to 40
Chemical Precipitation Effluent 3 to 10
Trickling Filter Effluent 3 to 10
Activated Sludge Effluent 2 to 8
Sand Filter Effluent 1 to 5
9.2.3 Upgrading Chlorine Contact Tanks
Improved disinfection can be achieved in existing chlorine contact tanks by:
1. Improving the mixing characteristics of the basin
2. Improving the flow pattern by elimination of short circuiting
3. Improving the chlorine diffuser system
9-5
-------
4. Lengthening the contact time
5. Increasing the chlorine dosage rate
6. Maintaining the optimum pH range
7. Upstream removal of ammonia nitrogen.
The technique to be employed will depend primarily on the disinfection standard to be met,
the characteristics of the plant effluent and physical characteristics of the plant itself.
Stringent disinfection standards required for protection of shellfish waters, recreational
waters and for various reuse purposes will often require nitrification of the ammonia
nitrogen present prior to chlorine addition unless extremely long postchlorination retention
basins are used to allow monitoring of plant effluents prior to discharge.
If less stringent disinfection is required, existing chlorination systems may be upgraded by
providing higher chlorine dosages, improved mixing and/or longer contact periods. In most
cases, improved mixing and longer contact times have proven more effective than increased
chlorine dosage. Effective contact time can be increased by improving the flow pattern
through the existing basin to eliminate short circuiting or by adding additional tankage.
The configuration of the contact tank may result in appreciable differences between the
actual and theoretical contact times. Model tests were made by the Metropolitan Sanitary
District of Greater Chicago (5) to evaluate the impact of different baffle designs on actual
detention time. The various baffle designs evaluated and the test results are shown on
Figure 9-3.
The data show that the baffle arrangement used in Scheme HA, using turning vanes,
provided substantially higher actual contact times than the other schemes tested. The
improper use of turning vanes, Scheme II, resulted in the lowest actual contact times. Based
on the findings of this study, the features of Scheme IIA were incorporated in the design of
the 330 mgd chlorination facility at the Chicago Calumet Sewage Treatment Plant. The
findings of this study saved the Metropolitan Sanitary District of Greater Chicago an
estimated $284,000 in construction costs and considerable reduction in annual costs due to
more efficient chlorine nrxing.
Rapid dispersement of chlorine at the addition point increases chlorine contact and
improves disinfection efficiency. Baffles can be designed to create turbulence at the chlorine
addition point and improve mixing. Baffled systems have the advantage of not requiring
mechanical equipment. Mechanical mixing or air agitation can be advantageously employed
where plant hydraulics will not allow the use of baffles, or where a portion of the existing
basin can be converted to a mixing chamber and the remainder of the basin and/or a long
outfall sewer can be used to provide the needed contact time.
9-6
-------
SCHEME
dD
FLOW. MGD
WATER DEPTH, FT
CONTACT TIME. MIN
MINIMUM
MEAN
MAXIMUM
82
14.
21
29.
36.
5
2
0
2
5
FIGURE 9-3
IMPACT OF CHLORINE TANK BAFFLE DESIGN ON ACTUAL DETENTION TIME (5)
SCHEME IA
/\
82 5
14 2
17 6
26,4
17 8
SCHEME IB
82.5
U. 2
15.3
23.9
34 8
SCHEME I I
82 5
M 2
15.2
20.5
31.9
SCHEME III
r
82.5
14.2
25.0
31. 1
39.5
-------
Studies by Kothandaraman and Lin (6) have shown that chlorinated wastewater can be
subjected to air agitation with no apparent loss of total chlorine residual. Their studies also
indicated a better bacterial kill with air agitation than without air agitation. In those
facilities where postaeration facilities are required to raise the DO content of the
wastewater, it may be feasible to use the chlorine contact tanks for both postaeration and
disinfection.
9.2.4 Chlorination Systems
The most common form of chlorine utilized in the field of wastewater treatment is liquid
chlorine. In a typical installation, liquid chlorine is delivered in pressurized containers.
Standard sizes for containers are 100-pound, 150-pound and 1-ton cylinders, and 16-ton,
30-ton and 55-ton tank cars. Normally, gaseous chlorine is released from the top of the
container and is drawn through the chlorinators by a vacuum. In larger plants where the
evaporation rate within the container cannot meet the demand, liquid chlorine is withdrawn
and evaporated in separate evaporators. The chlorine gas is then mixed with plant effluent
or potable water (small plants) to form a chlorine solution which is fed into the wastewater.
Such systems have proven to be highly reliable and economical. However, there are
ramifications that should be considered.
1. Liquid chlorine is a potentially dangerous material.
2. The demand for liquid chlorine has been increasing steadily, while the
construction of liquid chlorine manufacturing facilities has lagged.
3. Unless the above trend is corrected, it can be expected that shortages of liquid
chlorine will develop, and the cost will significantly increase.
In recent years, the utilization of sodium hypochlorite for disinfection has become
attractive. Sodium hypochlorite can either be delivered to the site in liquid form in 500 to
5,000 gallon tank cars or trucks, or manufactured on site. When delivered to the site, it is
normally sold at a concentration of 12 to 15 percent by weight of available chlorine. The
tank cars are unloaded into storage tanks, and the hypochlorite is fed to the wastewater
through chemical feeders. Sodium hypochlorite can be manufactured on site from salt or
from seawater. Generation units are now being produced by at least four major
manufacturers. While most of these units produce hypochlorite at concentrations of one
percent or less, one system produces hypochlorite at a concentration of about eight percent.
9.2.5 Ozonation
Based on the success of ozonation in the water treatment field, ozone has been considered
as a possible disinfectant for wastewater. Some of the advantages of ozonation are as
follows:
9-8
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1. Ozone is generated from air and its supply is dependent only on a source of
power. Ozone can also be produced from oxygen at significant energy savings, an
option which should be considered where on-site generation of oxygen is
practiced in conjunction with biological treatment.
2. Ozone may be a more efficient viricide than chlorine.
3. Ozonation can also be used as a tertiary treatment process for oxidation of
residual carbon compounds and for odor control.
4. The Maximum Allowable Concentration (MAC) of ozone in air as established by
the American Council of Governmental Industrial Hygienists is 0.1 ppm by
volume for continuous human exposure. The threshold odor of ozone is 0.01 to
0.02 ppm. This means that a person working near an ozone handling area should
be able to detect the presence of ozone at concentrations far below the MAC.
5. Ozonated effluents have not been shown to be toxic to the receiving water biota,
as have residual chlorine compounds such as chloramines and chlorinated
hydrocarbons. Ozone does not increase the dissolved solids concentration in the
effluent as does chlorine.
6. Ozonation of effluents increases the DO.
The predominant disadvantage of ozonation is the high capital and operating cost associated
with its generation. Further development of the process may lead to better economy which
would encourage greater use of ozonation for wastewater disinfection. .
Nebel, et al, (7) conducted pilot plant studies at Louisville, Kentucky, using ozone for
disinfection of activated sludge effluent. The wastewater contained a significant amount of
nonbiodegradable industrial organic wastes which exerted an above-normal ozone demand.
Nevertheless, the tests indicated that a dosage of 15 mg/1 and a contact time of 22 minutes
resulted in excellent destruction of fecal and total coliforms and fecal streptococci.
During 1973, two separate cost evaluations of alternative disinfection systems were made by
Metcalf & Eddy, Engineers in Boston. Both of the plants studied were large; one a 105-mgd
plant located on the eastern seaboard, and the other a 230-mgd plant located in the
midwest. The eastern plant study included cost comparisons of liquid chlorine, trucked-in
and on-site generated sodium hypochlorite and ozonation. The midwestern study compared
all of the above alternatives except ozonation. The results of the two studies are presented
in Table 9-2. The costs are shown as a ratio to the liquid chlorine cost and are a sum of the
annual operating costs and amortized capital costs.
9-9
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TABLE 9-2
RELATIVE TOTAL ANNUAL COSTS OF DISINFECTION ALTERNATIVES
Cost Ratio 1
Alternative
Liquid Chlorine
Sodium Hypochlorite, Trucked to Site
On-Site Generation
8% Sodium Hypochlorite
\% Sodium Hypochlorite
Ozone (Generated from Oxygen)
Eastern
Plant
105 mgd
1
2.0
1.2
1.8
3.0
Midwestern
Plant
230 mgd
1
2.4
1.4
2.1
Data used to compute cost ratios:
1. Cost of liquid chlorine
2. Cost of liquid sodium hypochlorite
3. Capital cost of on-site generator (1%)
4. Capital cost of on-site generator (8%)
5. Average ozone dosage
6. Average chlorine dosage
7. Amortization rate
}.49/ton
$0.18/gal
$4,200,000
$2,900,000
10 mg/1
16.6 mg/1
$75/ton
$0.20/gal
$1,116,000
$ 975,400
3 mg/1
40 yr. at6!/2% 15yr. at 7%
9.3 Odor Control
Wastewater treatment plants serving large municipalities are generally characterized by
extensive collection systems with correspondingly high detention times. For example, the
Washington, D. C. Pollution Control Plant serves areas as far as 25 to 30 miles away. This
type of situation often leads to odor problems during summer periods. Odor problems are
characteristically most critical during the plant's low flow periods (approximately 9 PM to 4
AM), due to increases in the sewer detention time.
9.3.1 Odor Generation and Characteristics
Odors from wastewater treatment plants can usually be attributed to three sources: septic
raw wastewater, overloaded secondary treatment facilities and sludge treatment practices.
Septicity in wastewaters is caused by the depletion of DO due to long residence in sewers
and the subsequent increase in anaerobic activity. As wastewater becomes septic, facultative
9-10
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and anaerobic bacteria flourish. These bacteria utilize nitrates and sulfates present in
wastewater as their oxygen source. The reduction of sulfate ions produces the highly
odorous gas, hydrogen sulfide. Other odorous gases which may be present are indole,
skatole, mercaptans, disulfides, volatile fatty acids and ammonia.
Increased summer temperature and extended sewer detention times can result in the rapid
buildup of hydrogen sulfide and carbon dioxide according to the following reactions (8):
S= + H20 + CO2
At a pH level below 8, the equilibrium shifts toward the formation of nonionized H2S and is
about 80 percent complete at pH 7. At pH 8 and above, most of the reduced sulfur exists in
solution as HS" and S~ ions (8). H^S is noticeable even in the cold when present in water to
the extent of 0.5 mg/1. When present to the extent of 1.0 mg/1, it becomes very offensive
(9).
Overloaded secondary treatment facilities are also a potential source of odor. If the air
supply to an activated sludge aeration tank is inadequate, odorous conditions usually
develop. It is also possible that a properly sized air supply system can strip odorous gases
from septic wastewater.
Odors associated with sludge treatment occur in thickening, digestion and sludge dewatering
facilities. Thickeners may receive both septic primary and secondary sludges. Gases from
well-operated digesters may contain small quantities of t^S, which are usually destroyed by
normal flaring of digester gas. The predominant odor in digested sludge is ammonia,
although traces of volatile organic acids may be present.
9.3.2 Odor Measurement
Odor data are generally qualitative rather than quantitative in nature. The two available
quantitative methods are the t^S determination and the Threshold Odor Number (TON).
The latter method is only semiquantitative in that determination of the TON is dependent
on the olfactory senses of the individual performing the analysis. This can be
de-personalized somewhat by using a panel to determine the TON value for a sample.
9-11
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9.3.3 Odor Control Methods Available
The various methods available for control of odor emanating from a wastewater treatment
plant are:
1. Changes in the operational procedures and new techniques
2. Chemical treatment or pretreatment, which might include chlorine, ozone, lime or
powered carbon
3. Collection and treatment of noxious gases.
9.3.3.1 Changes in Operational Procedures and New Techniques
Odors associated with septic wastewater are generally not amenable to solution through
operational changes within the treatment plant itself. These odors must be controlled
upstream of the plant through the use of aeration or chemical treatment methods. The
applicability of in-sewer aeration methods for reduction of odors and hydrogen sulfide
corrosion is discussed in the Process Design Manual for Sulfide Control in Sanitary Sewerage
Systems (10). Among the procedures that have been evaluated are U-tube installations (refer
to Section 8.2.1.4) and pure oxygen injection into force mains.
Many sludge odors in a plant are a direct result of an improperly operated or overloaded
anaerobic sludge digester. Improved temperature control and better mixing of digester
contents may alleviate the odor problem.
9.3.3.2 Chemical Treatment of Wastewater
Chlorination is probably the most widely used of the chemical treatment processes available
to control odors for the following reasons:
1. It is highly effective.
2. Many treatment operators have had experience in handling chlorine.
3. Chlorination facilities already exist at most plants for disinfection.
Chlorination is used for two purposes: to retard biological action which produces odors,
and to react chemically with odorous sulfur compounds, oxidizing them to innocuous sulfur
forms, usually free colloidal sulfur.
9-12
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1
Sulfide
mg/1
1.7
0.5
0.2
0.2
0.2
120
Sulfide
mg/1
1.7
0.7
0.2
0.2
0.2
Table 9-3 contains a summary of odor reduction data for a chlorinated raw domestic
wastewater (11). A prechlorination dosage of 10 mg/1 at maximum flow was recommended
for odor control. In addition, another incremental 5 mg/1 of chlorine capacity was
recommended as an adequate safety margin for peaks in sulfide levels or chlorine demand.
TABLE 9-3
EFFECT OF CHLORINE ON ODOR REDUCTION FOR A
RAW DOMESTIC WASTEWATER (11)
Detention Time in Minutes
Chlorine^
Dosage
mg/1
0
5
10
25
50
1 PH = 7.
Air Temperature = 85 deg F.
Ozone has been added to wastewaters for odor control with some favorable results. Because
of the extremely high reactivity of ozone, a much higher ozone demand is generally
exhibited by a wastewater than would be exhibited for chlorine. However, the use of pure
oxygen in activated sludge treatment may have an added benefit, since the oxygen gas could
provide the ozone generator with an economic source of oxygen. Due to the high cost of
ozone generation, the use of ozone for odor control may be limited (11).
Lime and powdered carbon have also been used in various applications for odor control. The
addition of lime to septic wastewater raises the pH. Since the solubility of H2S increases
with increasing pH, less ^S evolves, thereby decreasing the odor level. Powdered activated
carbon adsorbs odor-causing materials and thereby decreases the odor level. The results of a
laboratory odor study are presented in Table 9-4 (12). Concentrations of less than 10 mg/1
of powdered activated carbon were successful in providing significant odor reduction.
Additional design information for controlling odors is available in the Process Design Manual
for Sulfide Control (10).
9-13
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TABLE 9-4
EFFECT OF POWDERED ACTIVATED CARBON
ON ODOR REDUCTION (12)
Raw Domestic Activated Carbon Concentration of
Plant Wastewaterl Treated Effluent Activated Carbon
TON2 TON2 mg/1
Charlottesville, Va. 300 100 3.8
Hershey Estates, Pa. 140 100 5.0
Butler, Pa. 280 200 10.0
1 Sample temperature = 60 deg F.
2 Threshold Odor Number.
9.3.3.3 Collection and Treatment of Noxious Gases
The covering of odorous unit process facilities to localize odors is a method which can be
used to prevent odors from reaching the atmosphere. The major expenses of this method are
the covering of the units and collection and treatment of the evacuated gases. In cold
climates, covering units can lead to conditions of high humidity and indoor fog if proper
ventilation is not provided. Many municipal plants, e.g., Cedar Rapids, Iowa (13), and
Elmira, New York (13), are using low-cost, formed-in-place styrofoam domes on odorous
treatment units.
The treatment methods usually considered for evacuated gases include simple or catalytic
combustion, ozonation and chemical oxidation. Combustion methods require heating the
gases to approximately 800 to 900 deg F for catalytic combustion and approximately 1,300
to 1,400 deg F for simple combustion. Operating costs for these methods are primarily
determined by the amount of air to be heated. Ozonation costs, while somewhat affected by
the volume of gas collected, are primarily affected by the quantity of odorous materials to
be controlled.
9.3.4 Effects on Subsequent Units
A consideration in using chlorine for odor control is that the chlorine dosage should not
produce a high residual chlorine level which may in turn be detrimental to secondary
biological units. When using lime for odor control, consideration must also be given to
increased sludge production.
9-14
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9.3.5 Process Designs and Cost Estimates
A cost estimate has been prepared for two odor-control systems (chlorination and powdered
activated carbon) for 1, 3 and 5 mgd treatment plants. The capital costs are presented in
Table 9-5. The chemicals are added to raw wastewater before the downstream treatment
units.
TABLE 9-5
CAPITAL COSTS FOR ODOR CONTROL SYSTEMS
Plant Capital Costs for Odor Control Systems
Size Chlorination Powdered Activated Carbon
mgd 10mg/l 10mg/l
1 $46,000! $37,000
3 49,000! 52,000
5 51,0001 73,000
Smallest size commercially available chlorinator.
The chlorination system included a gas chlorinator capable of delivering 10 mg/1 of chlorine
during peak flow rates. A building, scale and other necessary appurtenances were included.
The powdered activated carbon system included a 15 day storage hopper and a volumetric
feeder capable of delivering 10 mg/1 at peak flow rates. In addition, a one-day capacity
slurry tank, pump, building and associated piping were included.
9.4 Other Uses of Chlorine
In the operation of wastewater treatment plants, chlorine has been found useful as an
upgrading technique. Some of the various applications of chlorine are as follows (4) (14)
(15):
1. Destruction or control of undesirable growths of algae and slime-forming bacteria
in pipelines and conduits
2. Control of filter flies, clogging and ponding in trickling filters (Chlorine applied
for approximately 8 hours to produce a residual of 1 to 2 mg/1 in the distributor
arm will generally unclog the filter. Residuals of 20 to 50 mg/1 will eliminate
ponding by causing the filter to unload all of its biological slime)
9-15
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3. Improvement in wastewater coagulation
4. Improvement in the separation of grease from wastewaters
5. Reduction of the immediate oxygen requirements of return activated sludge and
digester supernatant return
6. Chlorination of return activated sludge to control sludge bulking (A chlorine dose
of 1 to 10 mg/1 based on return sludge flow has been used, provided the chlorine
application point is located to allow for 2 to 3 minutes of mixing before being
discharged to the aeration basin; otherwise high "spot chlorine" concentrations
may damage the activated sludge biomass).
9.5 References
1. Unpublished U. S. EPA Data (October, 1973).
2. Chambers, C., Chlorination for Control of Bacteria and Viruses in Treatment Plant
Effluents. Journal Water Pollution Control Federation, 43, No. 2, pp. 228-241 (1971).
3. Browning, G. E. and McLaren, F. R., Experiences with Wastewater Disinfection in
California. Journal Water Pollution Control Federation, 39, No. 8, pp. 1351-1361
(1967).
4. Sewage Treatment Plant Design. Water Pollution Control Federation Manual of
Practice No. 8, Washington, D. C. (1959).
5. Louie, D., and Fohrman M., Hydraulic Model Studies of Chlorine Mixing and Contact
Chambers. Journal Water Pollution Control Federation, 40, No. 2, pp. 174-184 (1968).
6. Kothandaraman, V., and Lin, S. D., Air Agitation of Treatment Plant Effluents. Public
Works, 104, No. 8, pp. 65-68 (1973).
7. Nebel, C., Gottschling, R. D., Hutchison, R. L., McBride, T. J., Taylor, D. M., Pavoni,
J. L., Tittlebaum, M. E., Spencer, H. E., and Fleischman, M., Ozone Disinfection of
Industrial-Municipal Secondary Effluents. Journal Water Pollution Control Federation,
45, No. 12, pp. 2493-2507 (1973).
8. Sawyer, C., Chemistry for Sanitary Engineers. New York: McGraw-Hill Book
Company (1960).
9. Nordell, E., Water Treatment for Industrial and Other Uses. New York: Reinhold
Publishing Corporation (1961).
9-16
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10. Process Design Manual for Sulfide Control in Sanitary Sewerage Systems. U. S. EPA,
Office of Technology Transfer, Washington, D. C. (1974).
11. Roy F. Weston, Inc., Engineer's Preliminary Report Odor Control Studies Washington,
D. C. Water Pollution Control Plant. (December, 1967)
12. Aqua Nucharfor Odor Control in Waste Treatment, Covington, Virginia.
13. Dow Domes-Environmental Enclosures. Dow Chemical Company Bulletin, Midland,
Michigan (1968).
14. Fair, G., and Geyer, J., Water Supply and Waste-Water Disposal. New York: John
Wiley and Sons, Inc. (1954).
15. Operation of Wastewater Treatment Plants. Water Pollution Control Federation Manual
of Practice No. 1, Washington, D. C. (1970).
9-17
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-------
CHAPTER 10
SLUDGE THICKENING
10.1 Sludge Treatment
Raw sludge is normally unsuitable for disposal without prior treatment. Such treatment is
required to reduce the volume to be handled and to provide stabilization. Sludge treatment
is the most complex and costly aspect of wastewater treatment. Although the volume of
sludge produced is frequently less than one percent of the total volume of wastewater
treated, sludge treatment facilities can often account for 25 to 50 percent of the total
capital and operating costs for the entire plant (1).
10.1.1 Sludge Treatment Processes
The processes used for sludge treatment may be grouped into the following categories:
1. Thickening
2. Stabilization
3. Dewatering.
The important processes included in each of these categories are shown on Figure 10-1.
The separation of solids in either primary or secondary sedimentation tanks has two
objectives. These are the recovery of a high percentage of solids from the liquid portion as
required by effluent standards, and the concentration of these solids to reduce the required
capacity of the sludge processing units (2). These objectives are conflicting in that highest
effluent quality is achieved by immediate removal of settled solids, without allowing them
to accumulate in sedimentation tanks. Although primary sludges can often be effectively
thickened in the sedimentation tanks, it is preferable to remove secondary sludges quickly
and provide separate thickening units. Thickening of sludge to reduce its volume prior to
stabilization, dewatering, and ultimate disposal, in fact, generally provides the highest
benefit to cost ratio of all sludge handling processes(2). This chapter is directed to various
methods of sludge thickening.
Sludge stabilization processes are used to convert raw primary and secondary sludges to
inoffensive forms by reducing the organic material in the sludges or by otherwise rendering
them inert. This is especially essential when sludge is to be disposed of on land. Sludge
dewatering characteristics may also be improved by some methods of stabilization such as
thermal or lime conditioning. Where incineration is to be used for ultimate disposal, organic
10-1
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FIGURE 10-1
SLUDGE TREATMENT PROCESSES
THICKENING
STABILIZATION
DEWATERING
GRAVITY
THICKENING
1
3
ANAEROBIC
DIGESTION
1
3
VACUUM
FILTRATION
1
2
3
AIR
FLOTATION
2
CENTRIFUGATION
2
AEROBIC
DIGESTION
2
3
HEAT
TREATMENT
2
3
DRYING
BEDS
CENTRIFUGATION
1
2
3
1
3
NORMAL APPLICATION
1
CHEMICAL
CONDITIONING
1
2
3
PRIMARY SLUDGE ONLY
SECONDARY SLUDGE ONLY
COMBINED PRIMARY AND SECONDARY SLUDGES
PRESSURE
FILTRATION
1
2
3
10-2
-------
matter is required for self-supporting combustion and no prior organic reduction process
should be used. Sludge stabilization is considered in detail in Chapter 11.
Sludge dewatering is advantageous because it reduces the volume of sludge that will require
further processing and ultimate disposal. Where the sludge must be transported to a disposal
site, dewatering reduces the land requirements as well as handling costs. Dewatering prior to
incineration is necessary to remove excess water that otherwise would have to be removed in
the incinerator at the expense of auxiliary fuel. Sludge dewatering is the subject of
discussion in Chapter 12.
10.1.2 Sludge Processing Interrelationships
The various unit processes shown on Figure 10-1 may be combined in numerous different
sludge treatment configurations. Typical flow diagrams for several commonly used sludge
treatment systems are shown on Figures 10-2 and 10-3. The systems have been grouped into
two general categories, depending on whether or not biological sludge stabilization is
involved. Systems using biological stabilization are shown on Figure 10-2 and nonbiological
systems are shown on Figure 10-3.
The selection of the individual components of a sludge treatment system must not be made
independently of the other components to be used. Instead, a cost-effective system of
compatible sludge thickening, stabilization, and dewatering units must be selected on the
basis of the local situation. The method to be used for ultimate disposal of the sludge after
treatment may limit the selection of a sludge treatment system for a particular application.
The alternative process units that can be considered will depend on the type of wastewater
treatment, the sludge characteristics, and the space available for the treatment facilities.
Small plants are generally located in less densely populated areas where land availability is
not a primary constraint for either treatment facilities or ultimate sludge disposal and where
skilled operators may be difficult to obtain. In these instances a simple system such as is
shown on Figure 10-2(a) may be adequate. Stabilized sludge in this case is dried on sand
beds prior to ultimate disposal. Plants serving populations larger than 10,000 often find
mechanical devices more economical for sludge dewatering. A system using mechanical
sludge dewatering devices is shown on Figure 10-2(b). In this situation, waste activated
sludge is returned to the plant influent and the combined sludge is concentrated in the
primary clarifier without separate thickeners. Where separate thickening of waste activated
sludge is desirable prior to stabilization, a system such as shown on Figure 10-2(c) is
commonly used.
Nonbiological sludge processing systems are used mainly when:
1. The nature of the sludge makes it difficult to digest.
10-3
-------
FIGURE 10-2
TYPICAL SLUDGE PROCESSING SYSTEMS USING BIOLOGICAL STABILIZATION (3)
SLUDGE FROM SMALL
BIOLOGICAL
TREATMENT PLANT W
PRIMARY SEDIMENTA
(a)
ITHOUT
TION
AEROBIC OR
ANAEROBIC
SLUDGE
DIGESTION
DIGESTED
SLUDGE
SAND BEOS
1
DRIED SLUDGE TO
ULTIMATE DISPOSAL
SUPERNATANT TO
INFLUENT
PLANT
UNDERFLOW TO
PLANT INFLUENT
COMBINED WASTE
ACTIVATED AND
PRIMARY SLUDGE
FROM PRIMARY
SEDIMENTATION
ANAEROBIC
SLUDGE
DIGESTION
DIGESTED
SLUDGE
CHEMICAL
CONDITIONING
VACUUM FILTER
OR CENTRIFUGE
1
SLUDGE CAKE TO
ULTIMATE DISPOSAL
SUPERNATANT TO
PLANT INFLUENT
FILTRATE OR
CENTRATE TO
PLANT INFLUENT
u
PL
WASTE ACTIVATED
SLUDGE ^i
PRIMARY
SLUDGE
WERFLOW TO
(NT INFLUENT
t
FLOTATION
THICKENING
\
ANAEROBIC
SLUDGE
DIGESTION
VACUUM FILTER
OR CEMTRIFUGE
DIGESTED _
SLUDGE CHEMICAL ^ \ ^
^CONDITIONING *l J
1
SLUDGE CAKE
TO ULTIMATE
DISPOSAL
1 *
1 * ' FILTRATE OR CEHTRATE
1 ! SUPERNATANT TO T0 THICKENER
PLANT INFLUENT DIHUTIUCMICMT
GRAVITY
THICKENING
(OPTIONAL)
10-4
-------
FIGURE 10-3
TYPICAL SLUDGE PROCESSING SYSTEMS
USING NONBIOLOGICAL STABILIZATION (3)
VACUUM
FILTER
THICKENED
SLUDGE
(a)
HEAT
TREATMENT
r|
'
SLUDGE CAKE
-*. TO ULTIMATE
DISPOSAL
DECANT TO PLANT
INFLUENT OR
SEPARATE TREATMENT
FILTRATE TO THICKENER
OR PLANT INFLUENT
VACUUM
FILTER
THICKENED
SLUDGE
(b)
CHEMICAL
CONDITIONING
J
*\
EXHAUST GASES TO
SCRUBDER
t
FILTRATE TO THICKENER
OR PLANT INFLUENT
MULTIPLE-HEARTH
INCINERATOR
ASH TO ULTIMATE
DISPOSAL
-
THICKENED (c) CHEMICAL rrNTItinirr f
SLUOGE * CONDITIONING * CENTRIFUGE *
i
CENTRATE TO
THICKENER OR
PLANT INFLUENT
EXHAUST GASE
AND ASH
FLU ID IZED
BED
INCINERATOR
IXHAUST GASES T
i SCRUBBER
t
I
CYCLONE
* SEPARATOR
SH TO ULTIMATE
DISPOSAL
10-5
-------
2. A chemically or thermally conditioned sludge, dewatered by filtration or
centrifugation, is acceptable for land disposal and is the most economical
alternative.
3. Incineration is necessary for maximum sludge volume reduction thereby making it
uneconomical to provide anaerobic digestion, due to the loss of caloric value.
Nonbiological systems such as shown on Figure 10-3 may be used in these instances. Heat
treatment may be used to stabilize sludge organics prior to dewatering and ultimate disposal
as shown on Figure 10-3(a). Incineration systems such as shown on Figures 10-3(b) and
10-3(c) provide the highest degree of sludge volume and organic reduction. Although the
capital and operating costs of these units are high, the site requirements for disposing of the
residue are considerably reduced.
10.2 General Sludge Thickening Considerations
Sludge thickening is commonly used as the first step in a sludge treatment system to
separate more water from the sludge solids than can be accomplished in the wastewater
clarifiers. This reduces the required capacity and increases the efficiency of subsequent
sludge stabilization and dewatering processes that have higher unit costs than thickening.
The processes commonly used to thicken sludge are gravity thickening, air flotation
thickening, and centrifugation.
Gravity thickeners are used extensively since they are the most economical method of
thickening primary sludge or some blended primary and waste activated sludges. Gravity
thickeners are particularly applicable to smaller installations where the thickened sludge will
be anaerobically digested and dried on sand beds. Although waste activated sludge is
difficult to thicken, it often is effectively handled by mixing with primary solids prior to
gravity thickening. Where the weight of waste activated sludge is greater than 40 percent of
the total sludge weight, gravity thickening is not effective and separate thickening of waste
activated sludge by other methods should be considered.
Air flotation is a commonly used method of thickening for large quantities of waste
activated sludge. Under these circumstances, primary sludge can be thickened in the primary
clarifiers or by separate gravity thickening. Normally, air flotation thickening, in
conjunction with separate primary sludge thickening, will produce the highest overall sludge
concentration.
Centrifugation is a very adaptable process capable of thickening sludges with a wide range of
characteristics. Due to high operating, maintenance and power requirements, centrifugal
thickening is generally used only where spacial limitations or sludge characteristics make
other methods unsuitable.
10-6
-------
An important consideration in the use of any sludge processing unit is the effect of recycled
flows that are returned to the head of the plant. The quality of the overflow from well
operated gravity thickeners and the underflow from flotation thickeners is similar to that of
raw wastewater (1). However, return flows from thickeners are normally of sufficiently large
volume to result in recycled BOD and SS loads which can adversely affect overall plant
performance. With centrifugal thickening, high solids captures must be obtained to avoid
recycle of unsettleable fine solids to the head of the plant, which will adversely affect plant
effluent quality.
Thickening is an operation that produces a more concentrated and therefore more viscous
sludge. The more viscous sludge will result in higher friction losses than water or dilute
sludge, so it is important that the thickened sludge pumping and piping system be designed
to handle the sludge throughout the range of expected solids concentrations. Because the
consistency of the sludge may vary under different conditions, pumps must be able to
operate properly over a wide range of discharge pressures. The design of sludge piping
systems is complicated by the need for both large sizes to decrease the likelihood of
stoppages and to facilitate cleaning, and for high velocities to minimize the deposition of
grease and the settling of solids in the pipes. To minimize friction losses, sludge lines should
be as short as possible and should contain a minimum of bends and fittings.
10.3 Gravity Thickening
Gravity thickening is the most common process in use today for the concentration of sludge
prior to digestion and/or dewatering. Thickeners can contribute to the upgrading of sludge
handling facilities as follows:
1. Increase the hydraulic capacity of overloaded digesters or subsequent sludge
handling units.
2. Improve primary clarifier performance by providing continuous withdrawal of
sludge, thereby ensuring maximum removal of solids.
The process is simple and is the least expensive of the available thickening processes. The
reduction in size and improvement in efficiency of subsequent sludge handling processes
often can offset the cost of gravity thickening. The process also allows equalization and
blending of sludges, thereby improving the uniformity of feed solids to the following
processes. Existing gravity thickeners can be upgraded by providing continuous feed and
drawoff, by diluting the feed solids, and by chemical addition.
10.3.1 Process Considerations
Gravity thickening is characterized by zone settling. The four basic settling zones in a
thickener are:
10-7
-------
1. The clarification zone at the top containing the relatively clear supernatant.
2. The hindered settling zone where the suspension moves downward at a constant
rate and a layer of settled solids begins building from the bottom of the zone.
3. The transition zone characterized by a decreasing solids settling rate.
4. The compression zone where consolidation of sludge results solely from liquid
being forced upward around the solids.
To date, many attempts have been made to simulate zone settling in a batch settling test to
generate design information which would be applicable to a continuous unit. Various
theories have been developed for analyzing batch settling data and they have been reviewed
and discussed in the literature (4) (5) (6) (7). Most of the theories assume that the settling
velocity of sludge at a given concentration in a small batch cylinder is similar to the velocity
in prototype thickening units. However, it has also been recognized that other parameters
are involved, such as cylinder depth, cylinder diameter, mixing conditions, and sludge
characteristics. All have a definite influence on thickening performance. The cumulative
effect of these parameters is such that when batch settling test data are used for unit sizing,
the result is an oversized unit. For this reason, batch settling test results must be used with
caution (5). Edde et al (6) have developed a mathematical model from batch and full-scale
thickening data. The model facilitates determination of solids loading at a given sludge
blanket depth, initial feed solids, underflow concentration, and hydraulic loading. This
'technique is particularly useful for determining gravity thickener design parameters when
upgrading existing wastewater treatment plants.
10.3.2 Design Considerations
Both solids and hydraulic surface loadings must be considered when designing gravity
thickeners. Experience indicates that solids loading generally governs the design (8). Design
solids loadings and expected underflow concentrations are shown in Table 10-1.
The dry solids ratio of waste activated to primary sludge governs the acceptable solids
loading to be used in thickener design. As this ratio increases, the acceptable solids loading
decreases.
Most thickeners are operated at a hydraulic loading of 600 to 800 gpd/sq ft (9). Thickeners
with hydraulic loadings less than 400 gpd/sq ft have been found to produce odors (9). To
achieve hydraulic loadings in the acceptable range, secondary effluent is normally blended
with the combined waste sludge before feeding the resulting uniform diluted sludge to the
thickeners.
10-8
-------
TABLE 10-1
THICKENER DESIGN LOADINGS AND
UNDERFLOW CONCENTRATIONS
Underflow
Type of Sludge Concentration Average Solids Loading
%,TS psf/day
Primary 8-10 20
Primary and Trickling Filter 6-8 12
Primary and Waste Activated 3-6 8
(60:40 Weight Ratio)
As mentioned previously, solids loading is generally the controlling parameter and dictates
the required surface area of the thickener. For example, if a solids loading of 10 psf/day is
used for a combined primary and activated sludge and typical performance efficiencies are
desired, the calculated hydraulic loading will be in the order of 25 to 50 gpd/sq ft. Dilution
will be required to achieve the recommended 600 to 800 gpd/sq ft.
An important consideration in designing a thickener is to provide the capability to handle
peak conditions. Gravity thickeners are normally sized for average sludge production rates.
Peak sludge production rates often are adequately handled by storage available within the
system, either in the primary tanks or the thickener itself. However, a solids balance for the
system as it will be operated should be determined to indicate whether supplementary
storage or additional thickener capacity is required to accommodate the peak conditions.
Sufficient storage normally should be available within the system for the maximum
three-day plant solids loading.
Most continuous thickeners today are circular and designed with a side water depth of
approximately 10 feet. While sludge blanket depth is an important parameter, it has been
reported that underflow solids concentrations are independent of sludge blanket depths
greater than 3 feet (1). Increased sludge detention time in the thickener will result in
increased underflow solids concentration. A detention period of 24 hours has been
suggested as the time required to achieve maximum compaction (1). Sludge blanket depth
and detention time are closely interrelated. The sludge blanket depth may be varied with
fluctuation in solids production to achieve good compaction. During peak conditions the
detention time may have to be shortened to keep the sludge blanket depth sufficiently
below the overflow weirs to prevent excessive solids carry-over.
The performance of a gravity thickener depends a great deal upon the type of sludge to be
thickened. Generally, gravity thickening should not be used for thickening activated sludge
alone.
10-9
-------
In the design of gravity thickeners, it is important that operational flexibility be provided.
Such flexibility includes the ability to regulate the quantity of dilution water, adequate
sludge pumping capacity so that solids concentrations are not limited, continuous feed and
underflow pumping, protection devices against torque overload, and a sludge blanket
detection device.
10.3.3 Use of Gravity Thickening for Upgrading Existing Sludge Handling Facilities
Gravity thickening is often used prior to digestion processes. It can also be used as a
combined thickening and equalization process prior to sludge dewatering. Another
application is in areas where sludge hauling is utilized and there is a need to reduce the
volume of sludge to be. hauled. In all cases, gravity thickening will yield higher underflow
solids concentrations than obtainable with primary sedimentation, and the efficiency of
subsequent digestion and dewatering facilities will improve. Hence, gravity thickening
should always be considered in upgrading existing solids handling and dewatering facilities.
10.3.4 Upgrading Existing Gravity Thickeners
Improved thickening can be obtained by diluting the sludges to be thickened. It has been
reported that a feed solids concentration of 0.5 to 1.0 percent is optimum and that dilution
reduces the interference between the settling particles (1).
Torpey (10) used dilution for thickening combined primary and secondary sludges in the
development of the Densludge System. A feed sludge concentration of less than 1 percent
produced underflow concentrations of 11.2 and 6 percent for combined primary and
modified waste activated sludge, and combined primary and conventional waste activated
sludge, respectively. To obtain these dilute feed sludge concentrations, dilution water was
pumped from the primary or secondary clarifier and blended with the combined sludges
prior to thickening.
Thickening systems at New York City's Tallmans Island and Bowery Bay pollution control
facilities utilize the processes developed by Torpey and presently obtain 4 to 6 percent
underflow solids concentration with a yearly average of 4.5 percent (11). Both plants are
operating using a combination of the step and activated aeration processes. A similar design
with digested sludge recirculation at Bergen County, New Jersey, produces an underflow
concentration of 5.2 to 7.5 percent with a yearly average of 6.3 percent (12). In both cases,
the lower underflow concentrations occur during the summer months.
The improved thickening due to dilution can also be attributed to the fines that are washed
from the sludge and returned to the plant through the thickener overflow. Experiences at
Bergen County, New Jersey, show that, even with digested sludge recirculated to the
thickener, the overflow from the thickener does not appreciably affect the overall BOD
removal efficiency of the treatment plant (13).
10-10
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Similar experiences were reported at the Bowery Bay Pollution Control Plant. However, the
air requirements were increased from 0.28 to 0.31 cu ft/gal. (14) by the return of thickener
supernatant to the aeration tanks.
Experiences with the use of polyelectrolytes for upgrading gravity thickening have been
reported at Amarillo, Texas (15). The original plant was designed to handle 7.5 mgd, but
was receiving a flow of up to 10.5 mgd. A 55-foot diameter gravity thickener was being used
to thicken combined primary and waste activated sludge. However, bulking occurred in the
thickener due to the overloaded conditions and the low ratio of primary sludge to waste
activated sludge. In-plant recycling of solids resulted. To minimize the problem, only
activated sludge was thickened in the gravity thickener. Primary sludge was thickened in the
primary clarifiers. Polyelectrolyte addition was utilized in the thickener in an attempt to
improve sludge blanket control and to obtain maximum underflow solids concentration. It
was found that a cationic polyelectrolyte permitted a solids loading of 4.5 to 7.0 psf/day
while maintaining an underflow solids concentration of 2.6 percent, at a cost of $1.10 to
$3.64 per ton of dry solids. Polyelectrolyte was used here to successfully control the sludge
blanket height. This practice was continued until the plant was upgraded to 12 mgd.
For the expansion to 12 mgd at Amarillo, Texas, an existing 70-foot diameter final clarifier
was modified for thickening the waste activated sludge. The overflow from this thickener is
mixed with the primary sludge to dilute the feed sludge to the existing primary sludge
thickener (15). Operation of the waste activated sludge thickener showed that at a solids
loading of 2 to 3.5 psf/day, only a 2.4 percent underflow solids concentration could be
obtained. Polymer addition was tried once again to increase the underflow solids
concentration during a 142-day program, but proved unsuccessful.
In Chicago, the addition of polyelectrolyte at dosages of less than 10 Ib/ton dry solids
increased the solids loading by two to four times, but there was no improvement in
underflow solids concentration (16). From these two examples, it appears that
polyelectrolyte addition improves solids capture and reduces solids overflow, but has little
or no effect on improving solids underflow concentration.
10.3.5 Process Designs and Cost Estimates
The following example illustrates the use of gravity thickening before anaerobic digestion.
Existing anaerobic digesters were experiencing unstable operation due to the increased
amount of sludge generated by an increase in plant flow from 10 to 16 mgd. Operational
data from the overloaded plant without gravity thickening and the upgraded plant with
thickening are presented in Table 10-2. The volume of the combined primary and secondary
sludge was 100,000 gpd at 3 percent solids prior to upgrading. The increased sludge volume
due to the plant overloading decreased the detention time in the digesters from 17 to
11.25 days.
10-11
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TABLE 10-2
EXAMPLE OF
UPGRADING SLUDGE HANDLING FACILITIES
USING GRAVITY THICKENING
Description Overloaded Plant Upgraded Plant
Loadings to the Anaerobic Digester
Total Weight Primary and
Secondary Solids 25,100 Ib 25,100 Ib
Total Volume Primary and
Secondary Sludge 100,000 gpd (3%) 66,200 gpd (4.5%)
Digester Hydraulic Detention Time 11.25 days 17 days
To improve the operation of the existing digesters, it was decided to reduce the sludge
volume by gravity thickening. The flow diagram of the upgraded plant is shown on
Figure 10-4. The total volume of sludge discharged to the anaerobic digestion facility as a
result of gravity thickening was reduced from 100,000 gpd at 3 percent solids to 66,200 gpd
at 4.5 percent solids. The gravity thickener was designed using a mass loading of 10 psf/day.
A hydraulic overflow rate of 600 gpd/sq ft was achieved by recycling final effluent to the
mixing chamber ahead of the thickener as shown on Figure 10-4.
The capital costs for gravity thickening in this example are estimated at $305,000. These
costs include one gravity thickener, a mixing chamber, effluent recycle capacity, and an
allowance for appropriate connecting piping. They do not include the allowance for
engineering design, bonding, and construction supervision.
10.4 Air Flotation
The use of air flotation is limited primarily to thickening of sludges prior to dewatering.
Used in this way, the efficiency and/or capacity of the subsequent dewatering units can be
increased and the volume of supernatant from the subsequent digestion units can be
decreased. Existing air flotation thickening units can be upgraded by the optimization of
process variables, and by the utilization of polyelectrolytes.
Air flotation thickening is best applied to thickening waste activated sludge. With this
process, it is possible to thicken the sludge to 6 percent, while the maximum concentration
attainable by gravity thickening without chemical addition is 2 to 3 percent (1). The air
flotation process can also be applied to mixtures of primary and waste activated sludge. The
greater the ratio of primary sludge to waste activated sludge, the higher the permissible
10-12
-------
FIGURE 10-4
UPGRADING SLUDGE HANDLING FACILITIES USING GRAVITY THICKENING
RAW
WASTEWATER
o
I'
co
THICKENER
OVERFLOW
' PR
CLARIF
1
MIX
CHA
MARY
\
CATION! '
PRIMARY
SLUDGE
t
NG
MBER
*
AERATION
RETURN
WASTE
ACTIVATED
THICKENER D
/ GRAVITY ^
TH
i fc TD
/ FINAL \ EFFLUENT
'iCLARIF
V
SLUDGE
SLUDGE
LUTION WATER
CATION/
CKENED SLUDGE
n i CF?T i ny no
THICKENER
DEWATERING
-------
solids loading to the flotation unit. Due to the high operating costs, it is generally
recommended that air flotation be considered only for thickening waste activated sludge
(17).
10.4.1 Process and Design Considerations
The most commonly used type of air flotation unit is the dissolved air pressure flotation
unit. A schematic flow diagram for a typical unit is illustrated on Figure 10-5. In this unit,
the recycled flow is pressurized from 40 to 70 psig and then saturated with air in the
pressure tank. The pressurized effluent is then mixed with the influent sludge and
subsequently released into the flotation tank. The excess dissolved air then separates from
solution, which is now under atmospheric pressure, and the minute (average diameter
80 microns) rising gas bubbles attach themselves to particles which form the sludge blanket
(1). The thickened blanket is skimmed off and pumped to the downstream sludge handling
facilities while the subnatant is returned to the plant.
The following table is a summary of typical parameters used in the design of air flotation
thickening units.
TABLE 10-3
TYPICAL AIR FLOTATION DESIGN PARAMETERS
Without With
Parameter Polyelectrolyte Polyelectrolyte
Air Pressure, psig 40-70
Effluent Recycle Ratio, percent
of Influent Flow 30-150
Air to Solids Ratio, Ib air/lb solids 0.02
Solids Loading, psf/day 10 (average) 20-40
Polyelectrolyte Addition,
Ib/ton dry solids 0 5-10
Solids Capture, percent 70-90 90-96
Total Solids, percent
Unthickened 0.5-1.5
Thickened 4-6
A detailed discussion of the above parameters can be found in references (1) (17) (18) (19).
10-14
-------
FIGURE 10-5
SCHEMATIC OF AN AIR FLOTATION UNIT
SKIMMER MECHANISM
o
H-i
ca
RECYCLED _
SUBNATANT "
SBOTTOM COLLECTOR
j-Ai i
PRESSURE TANK
i
i
J
^
AIR
RECYCLED AUXILIARY
"RECYCLE
NFLUENT SLUDGE
-------
Bench-scale flotation units have been utilized for air flotation designs, but poor correlations
have generally been obtained with full-scale performance (17) (19). Therefore, pilot units
usually are recommended to determine optimum recycle rates, chemical requirements, and
general applicability of air flotation to sludge thickening.
Typical operating data for various air flotation units are presented in Table 10-4. Combined
primary and activated sludge produces a more concentrated float sludge than waste
activated sludge alone. Polyelectrolyte and/or chemical addition allows greater solids loading
and improves solids recovery without substantially increasing the float solids concentration.
However, the increased operating cost of the polyelectrolyte may largely offset the capital
cost savings resulting from the reduction of thickener area.
TABLE 10-4
AIR FLOTATION THICKENING PERFORMANCE DATA
Type of
Sludge
Waste Activated
Waste Activatedl
Waste Activated
Waste Activated
Waste Activated
Waste Activated
Waste Activated
Combined Primary and
Waste Activated
Combined Primary and
Waste Activated
Combined Primary and
Waste Activated
Solids
Loading
psf/day
12 to 18
24 to 48
13.9
7.1
19.8
26.2
28.8
Feed
Solids
percent
0.5 to 1.5
0.5 to 1.5
0.81
0.77
0.45
0.80
0.46
Float
Solids
percent
4.0 to 6.0
4.0 to 5.0
4.9
3.7
4.6
6.5
4.0
Solids
Recovery
percent
85 to 95
95 to 99
85
99
83
93
88
Reference
20
21
22
22
22
22
22
24 to 30
21
46.6
40.7
1.5 to 3.0 6.0 to 8.0
0.64
2.30
1.77
8.6
7.1
5.3
85 to 95
91
94
88
20
22
22
22
13 to 6 Ib polyelectrolyte/ton dry solids.
10-16
-------
10.4.2 Process Designs and Cost Estimates
The following example illustrates the use of prethickening by air flotation prior to anaerobic
digestion.
Existing anaerobic digesters were experiencing unstable operation due to the increased
volume of sludge produced by an increase in plant flow from 10 to 16 mgd. Operational
data from the overloaded and the upgraded plant is presented in Table 10-5. Prior to
upgrading, waste activated sludge was recycled to the primary clarifier. The volume of the
combined sludge was 100,000 gpd at 3 percent solids. The increased sludge volume due to
the plant overloading decreased the detention time in the digesters from 17 to 11.25 days.
To improve the operation of the existing digesters, it was necessary to reduce the sludge
volume to increase the digester detention time. To reduce sludge volume, thickening of the
waste activated sludge by air flotation was considered.
TABLE 10-5
EXAMPLE OF
UPGRADING SLUDGE HANDLING FACILITIES
USING AIR FLOTATION THICKENING
Description Overloaded Plant Upgraded Plant
Primary Solids, Ib 15,300 15,300
Primary Sludge Volume, gpd 30,600 (6%)
Waste Activated Solids, Ib 9,700 9,700
Waste Activated Sludge Volume, gpd 29,100 (4%)
Combined Solids, Ib 25,000 25,000
Combined Sludge Volume, gpd 100,000 (3%) 59,700 (5%)
Digester Hydraulic Detention
Time, days 11.25 18.8
The flow diagram for the upgraded plant is shown on Figure 10-6. As a result of separate
thickening of the waste activated sludge, it is expected that the primary sludge can be
concentrated to 6 percent in the primary clarifiers. The total volume of sludge discharged to
the anaerobic digestion facility due to the separate thickening of the waste activated sludge
and the improved solids concentrations in the primary clarifier is expected to be
59,700 gpd, compared to 100,000 gpd prior to upgrading. The air flotation system was
designed using an air pressure of 50 psig, an effluent recycle of 100 percent, and a solids
10-17
-------
FIGURE 10-6
UPGRADING SLUDGE HANDLING FACILITIES
USING AIR FLOTATION THICKENING
RAW
WASTEHATER
PRIMARY
CLARIFICATION
PRIMARY
SLUDGE
FLOTATION
SUBNATANT
AERATION
FINAL
CLARIFICATION
RETURN SLUDGE
AUXILIARY RECYCLE AIR
SLUDGE
STORAGE
TANK
AIR FLOTATION
UNIT
THICKENED SLUDGE
EFFLUENT
WASTE ACTIVATED
SLUDGE
POLYELECTROLYTE
DIGESTED SLUDGE
TO DEWATERING
10-18
-------
loading of 25 psf/day. It is anticipated that the polyelectrolyte dosage requirements will be
5 Ib/ton of dry solids.
The capital costs for air flotation thickening in this example are estimated at $183,000.
These costs include two air flotation units, a polyelectrolyte addition system, and
appropriate connecting piping. They do not include an allowance for engineering design,
bonding, and construction supervision.
Where separate thickening of primary and waste activated sludge is indicated, separate
gravity thickening of primary sludge in conjunction with air flotation thickening of waste
activated sludge will normally provide minimum thickening tank area. The following
example has been prepared to demonstrate this concept:
Combined
Gravity Separate
Thickening Thickening
Plant Flow, mgd 50
Primary Sludge, Ib/day 60,000
Waste Activated Sludge, Ib/day 40,000
Solids Loadings, psf/day
Combined Gravity Thickening 8
Primary Sludge Thickening 20
Waste Activated Sludge Thickening 25
Area Required, sq ft
Combined Thickening 12,500
Primary Sludge Thickening 3,000
Waste Activated Thickening - 1,600
Total 12,500 4,600
10.5 Centrifugation
The centrifugation process has been successfully used for many years in industry for
separating liquids of different density, for thickening slurries, and for removing solids.
Although the potential value of centrifuges for wastewater treatment has been recognized
for quite some time, only recently have they been installed for regular use in wastewater
treatment plants. The increasing use of centrifuges in the wastewater treatment field is the
result of recent improvements in centrifuge design, the availability of reliable performance
data, and the advantages of centrifuges in certain instances over other sludge processing
facilities.
10-19
-------
This discussion is limited to performance data for centrifuges used in sludge thickening and
the applicability of centrifuges for upgrading sludge thickening facilities. The types of
centrifuges used for sludge thickening and dewatering are discussed in Section 12.4.1 and
illustrated on Figure 12-2. The centrifuge process and design principles discussed in Sections
12.4.2 and 12.4.3 are applicable to both sludge thickening and sludge dewatering
applications.
10.5.1 Centrifuge Performance in Sludge Thickening
Due to high maintenance and power costs, centrifugal thickening is generally used only
where space limitations or sludge characteristics make other methods unsuitable.
Performance data for centrifuges used to thicken various types of sludge are shown in Table
10-6.
A disc centrifuge can thicken waste activated sludge to 4 to 5 percent solids without
polyelectrolytes. However, it is essential that this unit be preceded by coarse and fine
screening plus cyclonic grit removal to prevent clogging of the discharge nozzles.
A basket centrifuge can also thicken waste activated sludge. It can attain 8 to 10 percent
discharge solids, but thickens at a much lower rate than a disc centrifuge. No sludge
prescreening is required with the basket centrifuge.
A solid-bowl centrifuge can thicken waste activated sludge to 5 to 8 percent solids. The
addition of polyelectrolytes to the feed sludge significantly improves the solids feed rate
and/or captures efficiency in this type unit.
Centrifugal thickening of waste activated sludge was extensively studied at the Chicago
Sanitary District. Ettelt and Kennedy (16) evaluated both disc type and solid-bowl
centrifuges. The disc-type machine concentrated the activated sludge to about 7.0 percent at
6,000 rpm, but operational problems made its use impractical. Clogging of the sludge
discharge nozzles required repeated maintenance. Solid-bowl centrifuges, processing
activated sludge alone, thickened sludge from 6.6 to 7.5 percent.
A disc centrifuge has been successfully field tested for thickening waste activated sludge at
an eastern Pennsylvania community (26). Using a 30-inch centrifuge with a 150-hp motor
and 300-gpm feed rate, the disc centrifuge produced a 5 percent underflow with 90 percent
solids recovery. Since the plant did not have primary treatment, it was necessary to install a
screening device ahead of the centrifuge. The screening effectiveness was demonstrated in
that the nozzles of the centrifuge did not plug. The combination of effective screening and
patented recirculating system (allowing a larger nozzle size) was instrumental to the good
performance. Normally, however, disc centrifuges are not recommended where activated
sludge treatment has not been preceded by primary treatment.
10-20
-------
TABLE 10-6
CENTRIFUGAL THICKENING PERFORMANCE DATA
Type of Sludge
Waste Activated
Waste Activated
Waste Activated
(after Roughing Filter)
Waste Activated
(after Roughing Filter)
Waste Activated
Waste Activated
Waste Activated
Waste Activated
Centrifuge
Type
Disc
Disc
Disc
Disc
Basket
Solid-Bowl
Solid-Bowl
Solid-Bowl
Capacity
gpm
150
400
50-80
60-270
33-70
10-12
75-100
110-160
Feed Solids
percent
0.75-1.0
0.7
0.7
0.7
1.5
0.44-0.78
0.5-0.7
Underflow
Solids
percent
5-5.5
4.0
5-7
6.1
9-10
9-13
5-7
5-8
Solids
Recovery
percent
90+
80
93-87
97-80
90-70
90
90-80
65
85
90
95
Polymer
Requirement
Ib/ton
None
None
None
None
None
None
None
<5
5-10
10-15
Reference
23
23
23
23
23
24
25
25
-------
10.5.2 Use of Centrifugal Thickening for Upgrading Existing Sludge Handling
Facilities
The performance of centrifuges in various applications clearly indicates that centrifugal
thickening should be considered when the upgrading of solids handling facilities is required.
Centrifuges are a flexible upgrading device because of their applicability in the thickening of
various mixtures of sludges.
When used as a thickening device, a centrifuge can upgrade an overloaded anaerobic digester
by reducing the volume of feed sludge, thereby increasing digester detention time. In
addition, centrifuges can also be used to supplement existing overloaded gravity thickeners.
10.5.3 Process Designs and Cost Estimates
The following example is given to illustrate the upgrading of sludge handling facilities
through the use of centrifuges.
Existing anaerobic digestion facilities at an activated sludge plant became overloaded due to
an increase in plant flow from 10 to 16 mgd. As a result, the detention time in the digesters
was reduced to 11.25 days, thereby causing unstable operation of the digesters. Operational
data from the overloaded and the upgraded plant are presented in Table 10-7. Prior to
upgrading, waste activated sludge was recycled to the primary clarifier.
TABLE 10-7
EXAMPLE OF
UPGRADING SLUDGE HANDLING FACILITIES
USING CENTRIFUGAL THICKENING
Description Overloaded Plant Upgraded Plant
Primary Solids, Ib 15,300 15,300
Primary Sludge Volume, gpd 30,600 (6%)
Waste Activated Solids, Ib 9,700 9,700
Waste Activated Sludge Volume, gpd 23,300 (5%)
Combined Solids, Ib 25,000 25,000
Combined Sludge Volume, gpd 100,000 (3%) 53,900 (5.6%)
Digester Hydraulic Detention
Time, days 11.25 21
10-22
-------
To upgrade the digestion facilities, it was decided to thicken the waste activated sludge by
using a disc centrifuge, as shown on Figure 10-7. It was decided that polyelectrolyte
addition was not necessary. The volume of the waste activated sludge, 69,400 gpd at
1.7 percent solids (9,700 Ib dry solids/day) before centrifugal thickening was reduced to
23,300 gpd at 5 percent solids. Combination of the thickened waste activated sludge with
30,600 gpd of 6 percent primary sludge resulted in an increase in digester detention time to
approximately 21 days.
The capital cost for this upgrading procedure was estimated at $253,000. This cost includes
one operating and one standby disc centrifuge, in-line screens, sludge pumps, and
appurtenances, but does not include an allowance for engineering design, bonding, and
construction supervision.
10-23
-------
FIGURE 10-7
UPGRADING SLUDGE HANDLING USING CENTRIFUGAL THICKENING
RAW WASTEWATER
PRIMARY
CLARIFICATION
CENTRATE
SUPERNATANT
PRIMARY SCREENINGS AND GRIT
FINAL
CLARIFICATION
EFFLUENT
WASTE ACTIVATED
SLUDGE
IN-LINE
SCREENING
AND GRIT
REMOVAL
SLUDGE DEWATERING
10-24
-------
10.6 References
1. Burd, R.S., A Study of Sludge Handling and Disposal. Federal Water Pollution Control
Administration, Publication WP-20-4 (May, 1968).
2. Barnard, J. and Eckenfelder, W.W., Interrelationships in Sludge Separations. Included
in Water Quality Improvement by Physical and Chemical Processes, etc., by Gloyna, E.,
and Eckenfelder, W.W., Austin, Texas: University of Texas Press (1970).
3. Metcalf & Eddy, Inc., Wastewater Engineering: Collection, Treatment, Disposal.
McGraw-Hill, Inc. (1972).
4. Dick, Richard, and Ewing, Benjamin, Evaluation of Activated Sludge Thickening
Theories. Journal of the Sanitary Engineering Division, ASCE, 93, No. 4, pp. 9-29
(1967).
5. Vesilind, Arne, Design of Prototype Thickeners from Batch Settling Curves. Water and
Sewage Works, 115, No. 7, pp. 302-307 (1968).
6. Edde, Howard, and Eckenfelder, W., Theoretical Concept of Gravity Sludge
Thickening; Scale-Up Laboratory Units to Prototype Design. Journal Water Pollution
Control Federation, 40, No. 8, pp. 1486-1498 (1968).
7. Dick, R., Thickening. Included in Water Quality Improvement by Physical and
Chemical Processes, edited by Gloyna, E., and Eckenfelder, W.W., Austin,
Texas; University of Texas Press (1970).
8. Schroepfer, G.J., and Ziemke, N.R., Factors Affecting Thickening in Liquid Solids
Separation. National Institute of Health, Sanitary Engineering Report No. 156S
(March, 1964).
9. Sparr, A., and Grippi, V., Gravity Thickeners for Activated Sludge. Journal Water
Pollution Control Federation, 41, No. 11, pp. 1886-1904 (1969).
*
10. Torpey, W.N., Concentration of Combined Primary and Activated Sludges in Separate
Thickening Tanks. Journal of the Sanitary Engineering Division, ASCE, 80, No. 1,
pp. 1-17 (1954).
11. Private Communication with J. Donnellon, Department of Public Works, New York
City (December 10, 1970).
12. Zablatzky, H.R., and Baer, G.T., High Rate Digester Loadings. Journal Water Pollution
Control Federation, 43, No. 2, pp. 268-277 (1971).
10-25
-------
13. Private Communication with H. R. Zablatzky, Superintendent, Bergen County Sewer
Authority, Little Ferry, New Jersey (December 15, 1970).
14. Torpey, W.N., and Milbinger, N.R., Reduction of Digester Sludge Volume by
Controlled Recirculation. Journal Water Pollution Control Federation, 39, No. 9,
pp. 1464-1474 (1967).
15. Jordon, V.J., and Scherer, C.H., Gravity Thickening Techniques at a Water
Reclamation Plant. Journal Water Pollution Control Federation, 42,
No. 2, pp. 180-189 (1970).
16. Ettelt, G.A., and Kennedy, T., Research and Operational Experience in Sludge
Dewatering at Chicago. Journal Water Pollution Control Federation, 38, No. 2,
pp. 248-257 (1966).
17. Jones, Warren H., Sizing and Application of Dissolved Air Flotation Thickeners. Water
and Sewage Works, 115, No. 11, pp. R177-178 (1968).
18. Mulbarger, M.C., and Huffman, D., Mixed Liquor Solids Separation by Flotation.
Journal of the Sanitary Engineering Division, ASCE, 96, No. 4, pp. 861-871 (1970).
19. Ettelt, G.A., Activated Sludge Thickening by Dissolved Air Flotation. Proceedings
19th Industrial Waste Conference, Purdue University, pp. 210-244 (1964).
20. Katz, W.J., and Geinopolos, A., Sludge Thickening by Dissolved-Air Flotation. Journal
Water Pollution Control Federation, 39, No. 6, pp. 946-958 (1967).
21. Koogler, J.B., Operational Report of the Biddeford, Maine Sludge Disposal System.
Peapack, New Jersey: Komline-Sanderson Engineering Company (1966).
22. Katz, W.J., and Geinopolos, A., Concentration of Sewage Treatment Plant Sludges by
Thickening. Proceedings Tenth Sanitary Engineering Conference Waste Disposal
from Water and Wastewater Treatment Processes, University of Illinois (February 6-7,
1968).
23. Private Communication with George Patanaude, Philadelphia District Representative,
Sharpies-Stokes Division, Pennwalt Corporation, Wynnewood, Pennsylvania (October
27, 1970).
24. Eckenfelder, W.W., Industrial Water Pollution Control. New York: McGraw-Hill Book
Company (1966).
10-26
-------
25. Private Communication with Gene Guidi, Sales Manager, Environmental Control
Equipment, Bird Machine Company, South Walpole, Massachusetts (February 22,
1974).
26. Private Communication with Laurence Sheker, Resident Manager, Environmental
Equipment and Systems Division, Dorr-Oliver Incorporated, Camp Hill, Pennsylvania
(April 22, 1970).
10-27
-------
-------
CHAPTER 11
SLUDGE STABILIZATION
11.1 General
Sludge stabilization processes are used to convert raw wastewater sludges to inoffensive
forms by decreasing the organic content in the sludges or by otherwise rendering them inert.
This is especially essential when sludge is to be disposed of on land. The four major sludge
stabilization processes are anaerobic digestion, aerobic digestion, heat treatment and lime
stabilization. Principal design parameters and upgrading techniques for these processes are
discussed in this chapter. Additional information on the design of these processes is available
in the Process Design Manual for Sludge Treatment and Disposal (1).
A sludge stabilization process must be viewed as an integral component of the overall sludge
handling and treatment system. The selection and design of each of the components within
this system are interdependent. Important considerations in sludge processing
interrelationships are discussed in Section 10.1.2.
11.2 Anaerobic Digestion
Anaerobic digestion is one of the more commonly employed processes for sludge
stabilization in treatment plants over 1 mgd. The process biologically reduces the amount of
VSS that must be handled by subsequent dewatering and ultimate disposal operations,
renders the organic material nonputrescible and destroys a large number of pathogenic
organisms. However, since anaerobic digestion results in a breakdown of coarse solids into
finer particles, digested sludges are generally more difficult to dewater than undigested
sludges. Also, the high alkalinity of digested sludges may require a greater dosage of
conditioning chemicals than would be required with undigested sludges to achieve
comparable performance in the dewatering process.
The major gaseous end product of anaerobic digestion is methane, which is often used as a
source of fuel in wastewater treatment plants. The digested sludge is an excellent soil
conditioner and has found some utility for this purpose.
A major advantage associated with anaerobic digestion is its low energy requirement. Power
consumption is much less than that required for aerobic stabilization or heat treatment. The
methane produced is normally more than sufficient to generate the heat required to
maintain optimum digester temperatures and thus provides a surplus energy source that can
be used elsewhere in a wastewater treatment plant.
11-1
-------
11.2.1 Process Biochemistry
Anaerobic digestion of sludge is a complex biochemical process employing several groups of
anaerobic and facultative organisms. In general, the process can be considered to consist of
two steps. In the first step, facultative organisms called "acid formers" degrade the complex
organics of wastewater sludge to volatile organic acids, primarily acetic acid. In the second
step, these volatile acids are fermented to methane and carbon dioxide by a group of strict
anaerobes called "methane bacteria."
The more important of these two phases is the methane fermentation phase because:
1. The only mechanism of COD or BOD removal is the production of methane. Acid
production only solubilizes the complex organics; it does not accomplish
stabilization.
2. This step has been found to be the rate-limiting step in the reaction sequence.
The primary reason why the methane fermentation step is rate limiting is that the
reproduction rate for these organisms is quite low relative to that of other groups of
bacteria. For example, the doubling time of the acid formers is several hours while that of
the methane formers is, under ideal conditions, four days. Thus, even if a temporary
difficulty in the system arose, it would be much harder for the methane organisms to adjust
than for the acid formers. In addition, it has been found that the environmental conditions
required to maintain optimum performance by the methane organisms are much more
restrictive than for the acid-forming organisms. Consequently, most of the effort in
upgrading existing digesters should be expended to insure that the methane fermentation
step is carried out as efficiently as possible.
Improperly functioning digestion systems can be upgraded by applying procedures which
will make the systems more closely approach optimum performance conditions. Therefore,
before upgrading existing digesters, the environmental factors discussed below should be
investigated to determine if the units are operating at their full potential.
11.2.2 Environmental Conditions for Optimum Performance
Knowledge of the range of environmental conditions which favor optimum performance of
anaerobic digestion is not as extensive as desired. A summary of the state of knowledge is
given below.
11-2
-------
11.2.2.1 pH
Close pH control is required for this process because methane bacteria are extremely
sensitive to slight changes in pH. The usual pH range required is 6.6 to 7.4. In general, it is
wise to maintain the pH as close to 7.0 as possible.
In a properly operating anaerobic digester, pH is maintained naturally by a bicarbonate
buffer system due to the great quantity of carbon dioxide produced during methane
fermentation. The pH is a function of the bicarbonate alkalinity of the digesting liquor and
the fraction of 62 in the digester gas. Figure 11-1 prepared by McCarty (2) illustrates this
relationship. Because of the significance of pH control in digester operation, it is most
important that the dynamic nature of buffer destruction and formation in the digester be
understood. This process is reviewed in the following equations for simple carbohydrates
such as glucose. The equations mentioned are equally applicable to digestion of sludge.
acid formers
C6H1206 *- 3 CH3COOH
3 CH3COOH + 3 NH4HC03 ^3 CH3COONH4 + 3 H20 + 3 CO2
o /^TT ^^^IVTTT r,TT ^ methane bacteria
3 CH3COONH4 + 3H20 ^ 3 CH4 + 3 NH4HC03
The first equation represents the breakdown of glucose to acetic acid by acid-forming
bacteria. The acid is neutralized, as shown in the second equation, by the biocarbonate
buffer. If sufficient buffer is not present, the pH would drop, and the conversion of acetate
to methane, as shown in the third equation, would be inhibited. During the third reaction,
the buffer consumed in the second reaction is reformed.
A dynamic equilibrium between buffer formation and destruction is maintained when the
process is proceeding satisfactorily. However, when an upset occurs, it is usually the
methane bacteria rather than the acid formers which are adversely affected. Therefore, net
buffer consumption takes place, and the process is in danger of pH failure. When this occurs,
an external source of alkalinity such as lime must be added to maintain the pH in the proper
range.
Figure 11-1 indicates that the bicarbonate alkalinity should be maintained at a minimum
level of 1,000 mg/1 as CaCO3 to ensure adequate pH control. To determine the bicarbonate
alkalinity, both the volatile acid concentration and the total alkalinity must be measured.
Then,
Bicarbonate Alkalinity = Total Alkalinity - 0.8 Volatile Acids
The 0.8 factor in the above equation is required to convert the volatile acid units from mg/1
as acetic acid to mg/1 as CaCO3 the equivalent alkalinity unit.
11-3
-------
FIGURE 11-1
RELATIONSHIP BETWEEN pH AND BICARBONATE CONCENTRATION
GO
«c
C3
50
40
30 -
CNl
° 20
10
250
500
I
2500 5000 10,000
HCO: CONCENTRATION, MG/L AS CaC03
J
25,000
11-4
-------
It should be noted that in the second and third equations, ammonium bicarbonate was used
as the form of the alkalinity. This represents the situation in wastewater sludge digestion
where large concentrations of ammonium result from the destruction of protein. In fact, the
maximum value of the total alkalinity is set by the concentration of the ammonium ion.
The carbon dioxide generated in the methane fermentation will not form negatively-charged
bicarbonate (the buffer) unless an equivalent quantity of cation is present. This is provided
by the destruction of natural protein with the release of positively-charged ammonium. If a
cation is not present to force formation of bicarbonate buffer, self-regulation of pH in the
digestion process is not possible. In this case, alkaline material must be added continuously
to control the pH. For example, the anaerobic degradation of glucose, illustrated in the first
three equations, would require the addition of an external source of alkalinity. It is not
necessary that ammonium bicarbonate be utilized for this purpose; in fact, in terms of cost
and avoidance of potential toxicity, another bicarbonate salt might be favored.
In general, this difficulty will not be experienced in wastewater sludge digestion unless
either a high carbohydrate fraction from an industrial waste is present in the sludge or a very
thin sludge is being treated.
11.2.2.2 Temperature
The temperature response of methane bacteria is the same as other bacterial groups.
Although thermophilic methane bacteria exist, it is generally not economically feasible to
heat sludge to this temperature range. Thus, digestion of wastewater sludge is conducted in
the mesophilic range. The optimum temperature in this range is 35 deg C (95 deg F). More
important than maintenance of a particular temperature is maintenance of the chosen
temperature at a constant level. A temperature change of 1 or 2 deg C is sufficient to disturb
the dynamic balance between the acid formers and the methane formers. This will lead to an
upset because the acid formers will respond much more rapidly to changes in temperature
than will the methane bacteria. When heat exchangers are used to upgrade the performance
of existing digesters, care must be exercised to avoid wide temperature variations and
excessive temperature in the heat exchanger itself.
11.2.2.3 Nutrients
Speece and McCarty (3) have done the most definitive work on the macro-nutrient and
micro-nutrient requirements of methane bacteria. As these authors indicate, domestic
wastewater appears to contain all of the nutrients required by these organisms. However,
due to the uncertainty of the precise nutritional requirements of methane bacteria,
difficulty may be encountered in digestion when a considerable fraction of the wastewater is
of industrial origin.
11-5
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11.2.2.4 Toxic Materials
Substances which may be present in municipal sludges in concentration ranges which can
produce toxicity include heavy metals, sulfides, surface-active agents, light metals, and
certain organics. All of these can gain entrance to wastewater sludge from industrial sources.
In addition, light metal cations will enter sludge if an alkaline material is added to control
the pH. Several papers (4) (5) (6) review the best engineering data available on toxicity.
Reference should be made to these papers for complete information. General information
on some substances is shown in Table 11-1.
TABLE 11-1
CONCENTRATIONS WHICH WILL CAUSE A TOXIC SITUATION
IN ANAEROBIC DIGESTION OF MUNICIPAL SLUDGES
Substance Concentration
mg/1
Sulfides 200
Heavy Metals 1 > 1
Sodium 5,000- 8,000
Potassium 4,000 - 10,000
Calcium 2,000- 6,000
Magnesium 1,200 - 3,500
Ammonium 1,700- 4,000
Free Ammonia 150
1 Soluble.
It must be emphasized that the values in this table are only guides. If toxicity is suspected, a
thorough analysis of all the chemical constituents of the sludge must be made before
definite conclusions can be drawn. Potential solutions to toxicity problems, other than
elimination from the wastewater, should be evaluated in small-scale digesters of the type
used in laboratory investigations.
11.2.3 Anaerobic Digestion Systems
Prior to a discussion of procedures for upgrading the performance of anaerobic digestion
systems, a description of typical anaerobic digestion systems will be presented. Figure 11-2
illustrates the two types of digestion systems in use at present.
11-6
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FIGURE 11-2
ANAEROBIC DIGESTION SYSTEMS
GAS WITHDRAWAL
iti
INLET
LAYERx
SUPERNATANT
ACTIVE LAYER
STABILIZED
SOLIDS
REMOVAL
LOW-RATE DIGESTER
GAS WITHDRAWAL
INLET »r:
OUTLET
HIGH-RATE DIGESTER
OUTLET
11-7
-------
In a low-rate system, the tank is not mixed and, in some cases, is not heated. Sludge is added
at the top and withdrawn at the bottom. Stratification develops in the system due to a lack
of mixing. In general, this can be classified as a plug-flow system. Because of the lack of
mixing and consequent stratification, much of the digester volume is wasted, and many
operational problems result. In this type of digester, acidification sometimes takes place in
the top and middle layers while methane fermentation is confined to the lower layers. This
can lead to areas of low and high pH in the system, which restrict optimum biological
activity. Also, chemicals added for pH control are not dispersed throughout the tank, and
their effectiveness is limited. Grease breakdown is poor because the grease tends to float to
the top of the digester while the methane bacteria are confined to the lower levels. Methane
bacteria are removed with the digested sludge and are not recycled to the top, where they
are required. During progression from top to bottom of the digestion tank, the sludge is
compressed and gradually dewatered.
The high-rate system differs from the low-rate system in that the contents are well mixed,
either continuously or intermittently, and the digester is heated. This procedure avoids most
of the difficulties inherent in low-rate systems. Consequently, this system operates well at
lower HRT values and higher organic loading rates. Table 11-2 compares design criteria for
low-rate and high-rate digesters.
TABLE 11-2
TYPICAL DESIGN CRITERIA FOR LOW-RATE AND
HIGH-RATE ANAEROBIC DIGESTERS
Parameter Low-Rate High-Rate
Hydraulic Retention Time (HRT), days 30 to 60 15 to 20
VSS Solids Loading, pcf/day 0.04 to 0.1 0.15 to 0.30
Volume Criteria, cu ft/capita
Primary Sludge 2 to 3 1-1/3 to 2
Primary Sludge + Trickling Filter Sludge 4 to 5 2-2/3 to 3-1/3
Primary Sludge + Waste Activated Sludge 4 to 6 2-2/3 to 4
Various mixing systems have been successfully utilized in digesters. These include:
1. Single or multiple draft tubes through which sludge is circulated by a turbine
mixer set in the tube
2. Digester gas recirculation through diffusers in the base of the digester or drop
pipes.
11-8
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Specific design information for these systems can be obtained from the various
manufacturers.
Sludge heating is accomplished either (1) by circulating hot water through coils mounted on
the inner wall of the digestion tank, (2) by a heat exchanger furnished as part of the mixer
draft tube or (3) by circulating sludge through an external heat exchanger. The latter
method is preferred since it has been found that the coils inside the digestion tank are easily
caked with partially dried sludge. The sludge circulation rate to the external heat exchanger
depends on the efficiency of the heat exchanger and the temperature differential. Design of
a high-rate digestion system must include a heat balance to determine fuel requirements.
The WPCF Manual of Practice No. 8 (7) presents in detail the procedure for making such a
heat balance. This discussion includes valuable data on the fuel value of sludge gas and the
insulation characteristics of typical digester construction materials.
One difficulty with high-rate digestion is that the sludge leaving the digester is thinner than
the incoming sludge due to solids destruction. Secondary digesters are normally added to
many high-rate digestion systems to concentrate the sludge and provide supernatant
separation. These units are designed primarily as settling tanks, but also provide a source of
seed sludge in case of digester upset, and sludge storage where intermittent withdrawals for
dewatering or drying are made. If suitably equipped, they can also serve as standby
digesters.
11.2.4 Upgrading Existing Anaerobic Digestion Facilities
The first step in upgrading a digester is to evaluate process performance. This can be
accomplished with the aid of such tests as volatile acids and alkalinity and by a digester gas
analysis. Any sudden rise of volatile acids indicates that the system is out of biochemical
balance. A rise in the CC>2 fraction in the gas or a decrease in methane production per
pound of volatile solids added will also indicate upset. However, the volatile acid test is
more sensitive. When an upset occurs, an alkaline material must be added to maintain the
bicarbonate alkalinity above 1,000 mg/1 as CaCO%. An easily soluble bicarbonate salt, such
as NaHCO3, is best for this purpose. Care must be exercised not to exceed the level at which
the cation of the alkaline material will cause toxicity. If this is a potential problem, a
mixture of alkaline salts should be used. Kugelman and McCarty (5) have described methods
of preventing cation toxicity by adding appropriate quantities of cation antagonists.
Control of pH during an upset is only a temporary measure. The cause of the upset must be
located and eliminated. Sometimes this is easy. For example, heavy-metal toxicity can be
completely eliminated by precipitation of the metal in the digester as the sulfide (8). In
other cases, only exclusion of the toxin from the system will suffice.
11-9
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The major upgrading technique for low-rate digesters is conversion to high-rate digestion. To
maintain high-rate digestion, the following conditions are necessary:
1. Solids thickening to increase volatile solids loading and HRT
2. Complete mixing of digester contents
3. Uniform solids feed and withdrawal
4. Temperature control system capable of maintaining a uniform temperature at a
point within the range of 30 to 35 deg C
5. An average HRT of at least 15 days.
The principal techniques used for upgrading high-rate digesters are to increase feed solids
concentration, to provide a secondary digester for liquid-solids separation and to increase
HRT.
The relationship between thickening of solids, HRT and solids loading has been illustrated
by Sawyer (9) and is shown on Figure 11-3. This relationship points out the importance of
thickening the solids prior to digestion. Sludge thickening is discussed in Chapter 10.
Sections 10.3.5, 10.4.2 and 10.5.3 illustrate how the solids loading capacity of anaerobic
digesters can be increased by implementing sludge thickening techniques.
There are restrictions on the degree to which raw sludge can be thickened. These include the
difficulty of pumping thick sludge and the maintenance of adequate mixing in the digester.
Generally, sludge need not be thickened beyond about eight percent solids to optimize
digester volume utilization. If it is desired to thicken beyond this point, adequate studies of
the sludge flow characteristics must be made to evaluate pumping and mixing requirements.
The addition of digester mixing by itself can have a significant beneficial effect on digester
performance. At the City of Pontiac, Michigan (10), modifications were made to an existing
digester by adding a gas recirculation unit to improve mixing. Mixing inhibited scum
formation, improved heat transfer and provided a more stable digestion process. Many other
communities have had similar success (11).
Experiences at Chicago (12) have shown that digesters can be upgraded to operate at VSS
loadings of 0.2 pcf/day and at an HRT of 10 days. Complete mixing is necessary to operate
under these loading conditions.
Bergen County, New Jersey, is the best reported example of upgrading low-rate digesters to
high-rate digesters (13). In 1951, a 20 mgd activated sludge plant was constructed with four
1.3-million gallon digesters. In 1961, the capacity of the plant was increased to 50 mgd, and
11-10
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FIGURE 11-3
RELATIONSHIPS BETWEEN SLUDGE SOLIDS DIGESTER LOADINGS,
AND DETENTION TIME (9)
CO
CO
.40
.35
.30
.25
.20
. 15
. 10
.05
11-11
-------
the existing digesters were modified to high rate. In 1969, additional studies were conducted
on a full-scale basis, and it was found that two of the original digesters could handle the
entire loading from the existing 50 mgd activated sludge plant. On this basis, it has been
projected that the original four digesters will be able to handle the increased solids loading
when the plant is expanded to 75 mgd. It is significant to note that even at these higher
loadings, the anaerobic process was very stable and the efficiency of the process remained
the same.
Successful upgrading of existing digestion facilities at Bergen County was accomplished by
the following methods:
1. Completely mixing the digester contents
2. Prethickening the primary and secondary sludge to an average concentration of
6.3 percent (range 5.2 to 7.5 percent)
3. Increasing the applied solids loading from 0.22 pcf/day to 0.5 pcf/day as a result
of thickening
4. Decreasing the HRT from 22 to 10 days.
The loading to two-stage high-rate digestion systems may be increased by recirculating the
concentrated digested sludge from the secondary stage back to the primary stage, since this
effectively increases the SRT at the same digester HRT. As with activated sludge systems,
the limiting factor is the solids-liquid phase separation. Perhaps a degasification process can
be included between the primary and secondary tanks to aid in the separation, as is done
with the anaerobic contact process (14). When satisfactory phase separation is obtained in
the secondary digester, adequate SRT's can be maintained while decreasing HRT's to less
than 10 days (15). A more stable digestion process then results, due to higher populations of
methane bacteria in the primary stage and to the lessening of toxicity effects at longer
SRT's.
Digesters can also be upgraded by recirculating a portion of the digested sludge back to the
thickening units and mixing it with the incoming combined sludge and effluent recycle
water. This procedure has been reported by Torpey and Melbinger (16) with modified
aeration plants in New York City, and was originally adopted to improve the pumping
characteristics of highly concentrated digester feed sludge of 10 to 14 percent solids. It
appears that the thickening process was improved because the digested sludge was
incorporated into the pore spaces of the mixed primary and waste activated sludge, thereby
eliminating the typical gel structure produced by grease in the raw sludge. It was also found
that recycling digested sludge improved digester performance due to the seeding of the
combined sludge prior to digestion and because of the greater SRT thereby afforded.
Volatile solids reduction was also increased. A digested sludge recycle of 50 percent
11-12
-------
appeared to be optimum for the New York City plants, and a net volume reduction of
digested sludge from 197 cu ft/million gallons to 112 cu ft/million gallons was achieved.
It must be pointed out that conversion to high-rate digestion is not a cure-all, especially if
digested sludge is to be dewatered prior to final disposal. Methane production and volatile
solids reductions are approximately the same at high rate as at low rate, but indications are
that supernatant separation and dewatering of high-rate sludge are difficult (17). To obviate
this difficulty, Sawyer (9) has suggested that secondary digesters in high-rate digestion
systems be two to four times the capacity of high-rate primary digesters, to provide
adequate time for solids separation and conditioning. This technique has been reported at
Grand Rapids, Michigan (18), where the volatile solids loading to the primary digester is in
excess of 0.25 pcf/day. The ratio between secondary and primary digester capacity is 3.5:1.
In this case, secondary digester underflow solids exceed 10 percent and supernatant solids
average less than 2 percent of the raw solids load. This indicates that the economics of
decreasing the detention time in the primary digesters should be weighed against providing
the additional capacity in the secondary digesters when solids dewatering is required.
11.2.5 Anaerobic Digester Supernatant
Anaerobic digester supernatant constitutes a high strength recycle flow, as indicated by
Table 11-3. It imposes a significant incremental organic loading on biological treatment
processes. Separate supernatant treatment may be necessary in cases where available
treatment capacity is limited.
TABLE 11-3
TYPICAL PROPERTIES OF ANAEROBIC
DIGESTER SUPERNATANT (19)
Parameter Low Rate High Rate
mg/1 rcig/1
Total Solids 4,000 - 5,000 10,000 - 14,000
SS 2,000 - 3,000 4,000 - 6,000
BOD 2,000 - 3,500 6,000 - 9,000
VSS 650-3,000 2,400- 3,800
Alkalinity (MO) 1,000 - 2,400 1,900 - 2,700
H2S 70-90 190- 440
NH3-N 240-560 560- 620
PH 7.0-7.6 6.4- 7.2
11-13
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11.2.6 Impact of Alum and Iron Phosphorus Sludges on Anaerobic Digestion
The addition of alum or ferric chloride for phosphorus removal during biological treatment
substantially increases the weight of the solids generated by the treatment process (20) (21)
(22). However, in most instances, due to a higher sludge density, the volume of secondary
sludge after chemical addition does not increase in the same proportion. If the chemical is
added in the primary clarifier, the primary sludge concentration may be decreased. Piping,
pumping and process units should be sufficient in size and number to handle the change in
weight and/or volumes of sludges that result from chemical addition to either primary or
secondary treatment processes.
Many studies have indicated that neither ferric chlorine nor alum phosphorus sludges inhibit
anaerobic digestion (20) (21) (22) (23) (24). However, digester performance may be altered
due to changes in physical sludge characteristics or feed sludge alkalinity. Studies have also
indicated that chemically bound phosphorus is not released to the supernatant in the
anaerobic digestion process (20) (24) (25).
At Chapel Hill, North Carolina (26), it was reported that digester alkalinity was reduced in
the primary digester from 2,500 mg/1 to 1,500 mg/1 as CaCO3 after addition of alum to one
of two parallel trains (see Chapter 4), resulting in a need to add lime on one occasion.
Further, the secondary digester underflow concentration decreased from a normal range of
6 to 7 percent to 3.8 percent with a coincident increase in the supernatant SS concentration
from 1,000 mg/1 to 10,000 mg/1. In spite of these difficulties, the digestion process itself
produced a normal reduction in VSS throughout the entire alum treatment study.
11.2.7 Process Design and Cost Estimates
The following example illustrates the upgrading of a two-stage low-rate digestion facility to
a two-stage high-rate digestion system. This was required due to the increase in plant flow
from 1 to 3 mgd. A gravity thickener was added prior to digestion. Primary digester
performance was improved by adding gas mixing and installing external heat exchangers to
control the temperature more accurately. The comparison between existing and upgraded
design conditions is presented in Table 11-4. The flow diagram of the upgraded plant is
shown on Figure 11-4.
A mixing chamber ensures proper blending of sludges and effluent water prior to thickening.
With thickening, the solids concentration and the volatile solids loading to the primary
digester were increased from 2 to 5 percent and from 0.036 to 0.108 pcf/day, respectively.
Provision was made for recirculating digested sludge back to the thickener at a rate of
50 percent of the raw sludge feed.
11-14
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TABLE 11-4
UPGRADING AN EXISTING LOW-RATE DIGESTION SYSTEM USING
PRETHICKENING OF THE COMBINED SLUDGE AND
IMPROVEMENTS TO THE PRIMARY DIGESTER
Parameter
Plant Flow, mgd
Combined Sludge Characteristics
Volume, gpd
Solids Contribution, Ib/day
VSS, percent
Gravity Thickener
Number
Solids Loading, psf/day
Effluent Water Required to Dilute Sludge, gpd
Hydraulic Loading, gpd/sq ft
Thickened Sludge Volume, gpd
Digester
Number - Primary Digesters
Number - Secondary Digesters
Primary Digester Characteristics
Secondary Digester Characteristics
Digester Volume (Total), cu ft
Hydraulic Retention Time, days (Total)
days (Primary)
VSS Loading, pcf/day (Primary)
Original Design
1
9,100 (2%)
1,530
70
1
1
Limited heating and mixing
No heating or mixing
60,000
49.4
24.7
0.036
Upgraded Design
3
27,300 (2%)
4,590
70
1
10
248,100
600
10,920 (5%)
I
1
New gas mixing and improved heating
No heating or mixing
60,000
41.0
20.5
0.108
-------
FIGURE 11-^
UPGRADING AN EXISTING LOW-RATE DIGESTION SYSTEM
USING PRE-THICKENING OF THE COMBINED SLUDGE
AND IMPROVEMENTS TO THE PRIM ARY DIGESTER
RAW
*ATER w PRIMARY
THICKENER SUPERNATANTJ
k CLARIFIER
1
^ Ml)
i CHM
i
« C3
«t CD
r gf ^
JNG f1
KBER L
1^
^ AERATION
-------
The capital cost of this upgrading was estimated at $137,000, allocated as follows:
Thickener $ 74,000
Digester Renovation 63,000
Total $137,000
11.3 Aerobic Digestion
Aerobic digestion can be applied to the stabilization of primary sludges or combinations of
primary and secondary sludges. It has been indicated that aerobic digestion is competitive
with anaerobic digestion for activated sludge plants up to a size of at least 8 mgd (27).
Fewer operational problems are associated with aerobic digestion than with anaerobic
digestion. Hence, less laboratory control and daily maintenance are required. Also, the
dangers of gas explosions are eliminated because the only gaseous byproducts of aerobic
digestion are carbon dioxide and water vapor.
Typical concentrations of various constituents present in aerobic digester supernatant
liquors are listed in Table 11-5. Compared to anaerobic digester supernatant, the recycle
stream from aerobic digesters is relatively innocuous. The recycled SS, although high in
concentration, consists of stabilized solids which can be effectively removed in primary
settling and would have little impact on biological processes. The true organic loading,
attributable to this recycle, is represented by the soluble BOD and is similar in strength to
primary effluent.
TABLE 11-5
TYPICAL PROPERTIES OF AEROBIC
DIGESTER SUPERNATANT
Parameter Average Range Reference
mg/1 mg/1
SS 3,400 46 - 11,500 28
BOD 500 9- 1,700 28
Soluble BOD 51 4- 183 23
COD 2,600 228- 8,140 28
Alkalinty - 473 - 514 29
KjeldahlN 170 10- 400 28
Total P 98 19- 241 28
Soluble P 26 2.5- 64 28
pH 7.0 5.9- 7.7 28
11-17
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11.3.1 Process Considerations
Aerobic digestion is accomplished by aerating sludge until it is stable and relatively nuisance
free. In the aerobic digestion of sludges, two different forms of oxidation take place, as
shown in the following two reactions. First, a portion of the organic substrate in the sludge
is oxidized and the remainder is converted to cellular matter. Second, the cellular matter
produced is oxidized until only a relatively inert fraction remains.
(1) organic matter + C>2 - ^-cellular matter + CC>2 +
(2) cellular matter + 02 - *- digested sludge + CO2 +
The first reaction, biological oxidation, is most significant in the stabilization of sludges
with a high fraction of primary sludge. The second reaction, endogenous respiration,
predominates in the aerobic stabilization of waste activated sludge or trickling filter humus.
Since biological oxidation requires a higher oxygen transfer rate than endogenous
respiration, more air dissolution capacity must be provided for the stabilization of sludges
containing significant quantities of primary sludge.
Both pH and alkalinity are reduced in a properly functioning aerobic digester when
nitrification takes place. Nitrification lowers pH according to the following reaction:
1.5 02 » N02- + 2H+ + H20
The second step of nitrification is as follows:
- 0.5 Q Nitrobacter
Theoretically, 7.1 Ib of CaCO3 alkalinity are destroyed per pound of ammonia nitrogen
oxidized, since the two protons released neutralize one mole of CaCO.3 according to the
following reaction:
2H+ + CaC03 - «-H2CO3 + Ca2+
In batch-type aerobic digestion, it is possible that the pH may drop to a rather low value
(5.5±) at increased detention times. However, this does not seem to inhibit the process (30).
Batch-type aerobic digestion operating data shown in Table 11-6 indicate process
performance and the relationship between ammonia, nitrite, and nitrate nitrogen as a
function of detention time. Table 11-6 also indicates that in the normal temperature range
of 15 to 35 deg C, an increase in temperature increases the rate of VSS reduction.
11-18
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Detention
Time
days
5
10
30
60
TABLE 11-6
BATCH-TYPE AEROBIC SLUDGE DIGESTION
OPERATING DATA FOR MIXTURES OF PRIMARY
AND WASTE ACTIVATED SLUDGE (31)
Temperature
degC
15
15
15
15
vss
Reduction
percent
21
32
40.5
46
pJL
7.6
7.6
6.6
4.6
Alkalinity
mg/1
510
380
81
23
54
3.2
4.0
38
Trace
1.28
0.36
0.23
None
64
170
835
5
10
15
30
60
5
10
20
20
20
20
20
35
35
24
41
43
44
46
7.6
7.6
7.8
5.4
5.1
590
390
560
31
35
26
45
7.9
8.0
630
540
54
4.9
7.0
28
7.0
Trace
0.59
2.27
0.19
0.51
None
60
29
275
700
14
10.0
0.18
0.08
None
None
Pure oxygen rather than air may be used in the aerobic digestion process. Because air
contains only about 20 percent oxygen, the oxygen solubility in a system using air is only
1/5 that obtained when using pure oxygen. Consequently, pure oxygen may be used to
stabilize thicker sludges in which the high oxygen uptake rates cannot be satisfied with air
aeration.
A three-month plant scale study of aerobic digestion using pure oxygen in a covered system
was conducted at Speedway, Indiana (32). Aeration was accomplished in a covered 31,000
cu ft, four-stage reactor followed by a clarifier for decanting the sludge. The study was
divided into two phases, the first treating only waste activated sludge and the second
treating mixed primary and waste activated sludge. Much of the heat generated by biological
oxidation was retained within the covered system. This resulted in a significant increase in
sludge temperature and a corresponding increase in the rate of VSS destruction. The results
of this study are shown in Table 11-7. A positive DO level of 2 mg/1 was required for the
process to function well.
11-19
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TABLE 11-7
RESULTS OF HIGH-PURITY OXYGEN AEROBIC DIGESTION
SPEEDWAY, INDIANA (32)
Parameter Phase 1
Feed Sludge Type WAS*
Feed TSS, percent 2.14
VSS/TSS 0.77
Biodegradable VSS/TSS 0.45
Feed Temperature, deg C 19.5
Average Ambient Air Temperature, deg C 7.5
Sludge Temperature - Stage 4, deg C 33.0
HRT, days 16.3
Volatile Solids Loading Rate, pcf/day 0.064
VSS Reduction, percent 44
Biodegradable VSS Reduction, percent 94.6
1 Waste Activated Sludge.
2 Primary Sludge.
Phase 2
57% WAS + 43% PSL2
3.06
0.66
0.48
16.0
-2.2
31.0
11.6
0.109
43
89.6
Batch tests were conducted at the Metropolitan Denver Plant on a pure oxygen aerobic
digestion system using open tanks and a special oxygen diffuser system producing extremely
fine bubbles (33). These tests indicated that a 40 to 60 percent reduction in VSS can be
obtained with an HRT of ten days, when digesting waste activated sludge.
The overall cost effectiveness of using high-purity oxygen for aerobic digestion is greatly
enhanced if the wastewater treatment plant employs the high-purity oxygen activated sludge
process.
11.3.2 Design Basis
There are two methods of operating aerobic digestion tanks. One is on a continuous basis,
and the other is on an intermittent batch basis. The latter method is used most frequently in
smaller plants. Continuous digester operation requires provision for continuously decanting
the supernatant, but offers the advantages of equalizing air requirements and providing a
slow continuous supernatant return. In the batch operation, the sludge is supplied with air
in a completely mixed tank for a protracted period, followed by quiescent settling and
11-20
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decantation as new sludge is admitted. However, care must be taken in cold climates to
avoid creating a slush at prolonged detentions and subfreezing air temperatures.
In designing an aerobic digestion system, the characteristics of the raw sludge must be fully
identified. This can best be done with pilot plant studies. A small (55-gallon) aerobic
digestion pilot plant can be used for this purpose. Through such studies, the detention time,
oxygen requirements and optimum VSS reduction can be determined. Mathematical models
analyzing aerobic digestion kinetics have been reported by Reynolds (34) and by Smith, et
al (35).
Table 11-8 contains a summary of parameters used in the design of aerobic digestion units
for municipal wastewater sludges. When considering aeration requirements, recognition must
be given to the oxygen demand of the ammonia nitrogen present.
11.3.3 Use of Aerobic Digestion for Upgrading Sludge Handling Facilities
11.3.3.1 Use of Existing Facilities
As an upgrading technique, aerobic digestion can be carried out in existing unused tankage,
such as old Imhoff tanks or old clarifiers. If unusually shaped basins are used, attention
should be placed on ensuring that complete mixing will be achieved and that dead spots will
be eliminated. Potential dead spots can be filled and covered with concrete. Air-diffusion
systems are more easily adapted to unusual basin shapes than are surface aerators.
11.3.3.2 Supplemental Aerobic Digestion
Aerobic digestion can be used in conjunction with existing anaerobic digesters. In Monroe,
Wisconsin, and Corpus Christi, Texas, it has been found best to digest the primary sludge
anaerobically and the waste biological sludge aerobically (36). The advantage of this
segregation is that the primary sludge is not diluted by the waste biological sludge and the
anaerobically digested primary sludge subsequently filters better on a vacuum filter when it
does not contain the waste biological sludge.
11.3.3.3 Conversion of Anaerobic Digesters to Aerobic Digesters
If existing anaerobic digesters are overloaded and for some reason cannot be upgraded as
described in Section 11.2, they can be converted to aerobic digesters. Aerobic digesters yield
similar VSS reductions and are relatively odor free. Hence, this conversion may be
applicable to small overloaded plants in residential areas. Aerobic digestion will usually
require an increase in the blower capacity of the air supply system, which would
significantly increase the yearly operating cost. However, these increased costs could be
offset by savings in maintenance requirements. An alternative method of supplying the
additional air is by using mechanical surface aerators where tank geometry permits.
11-21
-------
TABLE 11-8
AEROBIC DIGESTION DESIGN PARAMETERS USING AIR
Parameter
Solids Retention Time
Volume Allowance, cu ft/capita
VSS Loading, pcf/day
Air Requirements
Diffuser System, cfm/1,000 cu ft
cfm/l,000cuft
Mechanical System, hp/1,000 cu ft
Minimum DO, mg/1
Temperature, deg C
VSS Reduction, percent
Tank Design
Power Requirement, BHP/10,000
Population Equivalent
Power Cost
$/yr/lb BOD Removed
$/yr/Capita
Value
10-151
15-202
3-4
0.024-0.14
20-351
>602
1.0-1.25
1.0-2.0
40-50
8-10
2.18
0.37
Remarks
Depending on temperature, type of sludge, etc.
Depending on temperature, type of sludge, etc.
Enough to keep the solids in suspension and
maintain a DO between 1-2 mg/1.
This level is governed by mixing requirements.
Most mechanical aerators in aerobic digesters
require bottom mixers for solids concentration
greater than 8,000 mg/1, especially if deep tanks
(> 12 feet) are used.
If sludge temperatures are lower than 15 deg C,
additional detention time should be provided so
that digestion will occur at the lower biological
reaction rates.
Aerobic digestion tanks are open and generally
require no special heat transfer equipment or
insulation. For small treatment systems (0.1 mgd),
the tank design should be flexible enough so that
the digester tank can also act as a sludge thickening
unit. If thickening is to be utilized in the aeration
tank, sock-type diffusers should be used to
minimize clogging.
These cost data are based upon three operational
plants in Pennsylvania and do not reflect current
escalation of energy costs.
Reference
36,37
29, 31, 38, 39
40,41,42
29, 36, 42, 43
1,17,35,40
1, 35, 36, 43
40,44
38,40
17
35
40
1 Waste activated sludge alone.
2 Primary and waste activated sludge, or primary sludge alone.
11-22
-------
When existing anaerobic digesters are converted to aerobic digesters, the impact of retaining
or removing the digester covers on sludge temperature and the temperature of the air under
the covers, accessibility of diffused air equipment, generation of objectionable odors during
warm weather periods, oxygen demand and air supply capacity must be considered.
Experience at one midwestern city indicated that the combination of an air-diffusion system
and a covered aerobic digester (converted from an existing anaerobic digester) increased the
air temperature inside the digester to 135 deg F and the sludge temperature to 97 deg F
(45). At these high temperatures, objectionable odors were produced due to high
concentrations of organic sulfides in the sludge and the low digester DO. Under these
conditions the sulfur-bearing compounds were not completely oxidized resulting in the
release of f^S to the atmosphere. The increased oxygen demand of the sludge that occurred
at the unanticipated high temperatures also reduced the ability of the air supply system to
maintain the desired DO levels.
11.3.4 Process Design and Cost Estimates
The following example illustrates upgrading sludge handling facilities with aerobic digestion.
Due to continued operational problems with an existing anaerobic digestion facility, a
community decided to convert the existing anaerobic digesters to aerobic digesters by the
installation of blower and diffused air equipment. The waste activated sludge from the 4
mgd activated sludge plant is settled along with raw wastewater in the primary clarifiers and
pumped directly to the digestion system at an average solids concentration of 3 percent. The
upgraded facility is shown on Figure 11-5.
In the upgraded system, the two aerobic digesters will be operated continuously at a total
detention time of 37.4 days. This detention time is in excess of the 15 to 20 days required
and will allow the aerobic digestion process to handle increased solids loading in the future.
A summary of the design data for the upgraded facility is presented in Table 11-9.
The estimated total capital costs for the upgrading are $203,000, broken down as follows:
Air System $157,000
Renovation to Existing Tank 46,000
Total $203,000
11.4 Heat Treatment of Sludge
The heat treatment process involves heating sludge for short periods of time under pressure.
It is essentially a conditioning process which prepares sludge for dewatering without the use
of chemicals. In addition, the sludge is sterilized and generally rendered inoffensive.
11-23
-------
COMBINED PRIMARY AND
WASTE BIOLOGICAL SLUDGE
FIGURE 11-5
CONVERSION OF ANAEROBIC TO AEROBIC DIGESTION
DIGESTER 1
(EXISTING STRUCTURE)
AIR HEADER
1 r
EXISTING SLUDGE
PUMPS
DIGESTER 2
(EXISTING STRUCTURE)
STABILIZED SLUDGE
TO DENATERING
FACILITIES
SUPERNATANT
-------
TABLE 11-9
AEROBIC DIGESTER UPGRADING DESIGN PARAMETERS
Description Value
Plant Design Flow, mgd 4
Number of Digesters 2
Volume (each), cu ft 75,000
Total Solids Added to Aerobic Digesters, Ib/day 7,500
Volatile Solids Added to Aerobic Digesters (70 percent volatile), Ib/day 5,250
Sludge Volume (3 percent solids), gpd 30,000
VSS Loading (each) pcf/day 0.035
Hydraulic Detention Time (total), days 37.4
Aeration Requirements (each), cfm 4,500
1 (60 cfm/1,000 cu ft).
When sludge is conditioned by heat treatment, thermal activity releases the bound water and
results in coagulation of solids. In addition, hydrolysis of proteinaceous material occurs
resulting in cell destruction and ammonia nitrogen release. Approximately 30 to 40 percent
of the VSS are solubilized resulting in a very high strength recycle liquor.
Two variations exist for heat treatment. In both systems, sludge is passed through a heat
exchanger into a reactor vessel, where steam is injected directly into the sludge to bring the
temperature to within the range of 144 to 200 deg C, under a pressure of 150 to 200 psig.
In one variation air is also injected into the reactor vessel with the sludge. The detention
time in the reactor is approximately 30 minutes. After heat treatment, the sludge passes
back through the heat exchanger to recover heat, and then is discharged to a
thickener-decant tank. The thickened sludge may be dewatered by vacuum filtration or
centrifugation to a solids content of 30 to 45 percent.
The supernatant and filtrate recycle liquor contains high concentrations of short-chain water
soluble organic compounds and ammonia nitrogen. The strength of the recycle is much
higher than that from anaerobic or aerobic digestion, but is generally amenable to biological
treatment. This flow should be returned to the plant for treatment but careful consideration
must be given to the effects of the increased organic and ammonia nitrogen loads.
Otherwise, the liquor must be treated separately and then returned to the plant. A complete
discussion of the heat treatment process is presented in the Process Design Manual for
11-25
-------
Sludge Treatment and Disposal (1). Characteristics of filtrate for a heat treated sludge with
and without air addition are listed in Table 11-10. The results presented are from a pilot
study on high-purity oxygen activated sludge with a feed concentration of 2.68 percent
from the Batavia, New York, Wastewater Treatment Plant.
TABLE 11-10
CHARACTERISTICS OF HEAT TREATED SLUDGE FILTRATE
AT BATAVIA (46)
Parameters With Air Injection Without Air Injection
mg/1 mg/l
COD 17,000 15,900
BOD 7,210 7,650
TSS 8,800 11,400
VSS 7,800 10,200
Total P 250 140
Total N 1,610 1,510
NH3-N 740 540
pH 4.7 6.4
The advantages claimed for the heat treatment process are:
1. Excellent dewatering characteristics of treated sludge without chemical
conditioning
2. Innocuous sludge suitable for ultimate disposal by a variety of methods
3. Few nuisance problems
4. Suitable for many types of sludge which cannot be stabilized biologically
5. Reduction in subsequent incineration energy requirements
6. Reduction in size of subsequent vacuum filters and incinerators.
11-26
-------
Disadvantages claimed for the heat treatment process are:
1. High construction and operation
2. Specialized supervision and maintenance requirements due to high temperatures
and pressures
3. High recycle organic and ammonia nitrogen loads
4. Expensive material costs to prevent corrosion.
11.5 Lime Stabilization of Sludge
Land disposal of raw sludge is objectionable primarily because the sludge contains a large
quantity of pathogenic microorganisms and heavy metals. Furthermore, biological
conversion of the sludge cake into an inert material can create odor problems. Lime
stabilization of raw sludge can improve its acceptability for land disposal.
Various investigators have reported that the addition of lime to raw or digested sludges to
pH ranges of 10.2 to 12.5 has effectively reduced the number of pathogenic organisms
present (47) (48) (49). Current U. S. EPA sponsored work indicates that the pH should be
increased to 12.0 for more effective disinfection (1). Farrell, et al, have investigated the
bactericidal effect of lime on chemically precipitated primary sludges. Their findings are
presented in Table 11-11.
TABLE 11-11
BACTERICIDAL EFFECT OF LIME ADDITION TO
CHEMICALLY PRECIPITATED SLUDGES (50)
Bacterial Count (organisms/liter of sludge)
Alum-Primary
Limed Alum-Primary
Ferric-Primary
Limed Ferric-Primary
Salmonella
Species
110
None detected
> 24,000
None detected
Pseudomonas
aeruginosa
1,300
None detected
610
None detected
Total Aerobic
Count
41 x 108
5.0 xlO8
190 x 108
0.29 x 108
11-27
-------
Lime addition to raw sludge does not render it permanently stable because the pH
eventually drops and surviving organisms, or organisms that recontaminate the sludge, can
create nuisance conditions (50). Disposal in sanitary landfills of dewatered lime stabilized
raw sludge has been suggested as a satisfactory procedure. In such cases, the sludge should
be deposited in thin layers and covered daily.
11.6 References
1. Process Design Manual for Sludge Treatment and Disposal. U. S. EPA, Office of
Technology Transfer, Washington, D. C. (1974).
2. McCarty, P.L., Anaerobic Waste Treatment Fundamentals. Public Works, 95, No. 9, pp.
107-112(1964).
3. Speece, R.L., and McCarty, P.L., Nutrient Requirements and Biological Solids
Accumulation in Anaerobic Digestion. Proceedings of the International Conference on
Water Pollution Research, Pergamon Press (1962).
4. Kugelman, I.J., and Chin, K.K., Toxicity Synergism and Antagonism in Anaerobic
Waste Treatment Processes. Presented before Division of Air, Water and Waste
Chemistry, American Chemical Society, Houston, Texas (February, 1970).
5. Kugelman, I.J., and McCarty, P.L., Cation Toxicity and Stimulation in Anaerobic
Waste Treatment. Journal Water Pollution Control Federation, 37, No. 1, pp. 97-115
(1965).
6. Lawrence, A.W., Kugelman, I.J., and McCarty, P.L., Ion Effects in Anaerobic
Digestion. Technical Report No. 33, Department of Civil Engineering, Stanford
University (March, 1964).
7. Sewage Treatment Plant Design. Water Pollution Control Federation Manual of
Practice No. 8, Washington, D.C. (1959).
8. Lawrence, A.W., and McCarty, P.L., The Role of Sulfide in Preventing Heavy Metal
Toxicity in Anaerobic Treatment. Journal Water Pollution Control Federation, 37,
No. 3, pp. 392-409 (1965).
9. Sawyer, C., Anaerobic Units. Proceedings of a Symposium on Advances in Sewage
Treatment Design, Metropolitan Section-Sanitary Engineering Division, ASCE, New
York (1961).
10. Meyers, H.V., Improved Digester Performance through Mixing. Journal Water Pollution
Control Federation, 33, No. 11, pp. 1,185-1,187 (1961).
11-28
-------
11. Langford, L.L., P.F.T. - Pearth Multipoint Gas Recirculation. Water and Sewage
Works, 108, No. 10, pp. 382-383 (1962).
12. Lynam, B., et al, Start-Up and Operation of Two High-Rate Digestion Systems. Journal
Water Pollution Control Federation, 39, No. 4, pp. 518-535 (1967).
13. Zablatzky, H., and Peterson, S., Anaerobic Digestion Failures. Journal Water Pollution
Control Federation, 40, No. 4, pp. 581-585 (1968).
14. Schroepfer, G.J., et al, The Anaerobic Contact Process as Applied to Packing House
Wastes. Sewage and Industrial Wastes, 27, No. 4, pp. 460-486 (1955).
15. Dague, R., Application of Digestion Theory to Digester Control. Journal Water
Pollution Control Federation, 40, No. 12, pp. 2,021-2,031 (1968).
16. Torpey, W., and Melbinger, N., Reduction of Digested Sludge Volume by Controlled
Recirculation. Journal Water Pollution Control Federation, 39, No. 9, pp. 1,464-1,474
(1967).
17. Burd, R.S., A Study of Sludge Handling and Disposal. Federal Water Pollution Control
Administration, Publication WP-20-4 (May, 1968).
18. Voshel, D., Gas Recirculation and CRP Operation. Wastes Engineering, 34, No. 9,
pp. 452-455 (1963).
19. Malina, J.F., Jr. and DiFilippo, J., Treatment of Supernatants and Liquids Associated
with Sludge Treatment. Water and Sewage Wastes, pp. R-30-R-38 (1971).
20. The Dow Chemical Company, Application of Chemical Precipitation Phosphorus
Removal at the Cleveland Westerly Wastewater Treatment Plant. Report prepared for
the City of Cleveland, Ohio, by the Dow Chemical Company (April, 1970).
21. Derrington, R.E., Stevens, D., and Laughlin, J.E., Enhancing Trickling Filter
Performance by Chemical Precipitation. U. S. EPA, Grant No. S800685, Project
No. 11010 EGL (August, 1973).
22. Long, D.A., Nesbitt, J.B., and Kountz, R.R., Soluble Phosphorus Removal in the
Activated Sludge Process. Report for the Water Quality Office, U. S. EPA, Project
No. 17010 EIP (August, 1971).
23. Thompson, J.C., Removal of Phosphorus at a Primary Wastewater Treatment Plant.
Paper presented at the Spring Meeting, New England Water Pollution Control
Association, Stratton, Vermont (June 7, 1972).
11-29
-------
24. Johnson, E.L., and Beeghly, J.H., and Wukasch, R.F., Phosphorus Removal at Benton
Harbor St. Joseph, Michigan. Report prepared for Benton Harbor St. Joseph,
Michigan, Joint Board of Commissioners by the Dow Chemical Company (1968).
25. Zenz, D.R., and Pivnicka, J.R., Effective Phosphorus Removal by the Addition of
Alum to the Activated Sludge Process. Proceedings of the 24th Industrial Waste
Conference, Purdue University, Lafayette, pp. 273-301 (May, 1969).
26. Brown, J.C., Little, L.W., Francisco, D.E., and Lamb, J.C., Methods for Improvement
of Trickling Filter Plant Performance, Part II, Alum Treatment Studies, U. S. EPA,
Contract No. 14-12-505, University of North Carolina, Chapel Hill, N. C. (1974).
27. Smith, A.R., Aerobic Digestion Gains Favor. Water and Wastes Engineering, 8, No. 2,
pp. 24-25 (1971).
28. Ahlberg, N.R., and Boyko, B.I., Evaluation and Design of Aerobic Digesters. Journal
Water Pollution Control Federation, 44, No. 4, pp. 634-643 (1972).
29. Malina, J.F., and Burton, H.M., Aerobic Stabilization of Primary Waste water Sludge.
Proceedings-19th Industrial Waste Conference, Purdue University, Lafayette, Indiana,
pp. 716-723 (May, 1964).
30. Metcalf & Eddy, Inc., Wastewater Engineering: Collection, Treatment, Disposal.
McGraw-Hill Book Company, New York, page 612 (1972).
31. Jaworski, N., et al, Aerobic Sludge Digestion. Presented at the Conference on
Biological Waste Treatment, Manhattan College, N. Y., N. Y. (April 20-22, I960).
32. Smith, J.E., Jr., Young, K.W., and Dean, R.B., Biological Oxidation and Disinfection of
Sludge. Prepublication copy (1973).
33. Cohen, D.B., and Puntenney, J.L., Metro Denver's Experiences with Large Scale
Aerobic Digestion of Waste Activated Sludge. Presented at the 46th Annual Conference
of the Water Pollution Control Federation, Cleveland, Ohio (October , 1973).
34. Reynolds, T., Aerobic Digestion of Waste Activated Sludge. Water and Sewage Works,
114, No. 22, pp. 37-42 (1967).
35. Smith, R., Eilers, R.G., and Hall, E.D., A Mathematical Model for Aerobic Digestion.
U. S. EPA, Office of Research and Monitoring, Advanced Waste Treatment Research
Laboratory, Cincinnati, Ohio (February, 1973).
11-30
-------
36. Dreier, D.E., Aerobic Digestion of Solids. Proceedings-18th Industrial Waste
Conference, Purdue University, Lafayette, Indiana, pp. 123-139 (1963).
37. Loehr, R.C., Aerobic Digestion: Factors Affecting Design. Water and Sewage Works,
Reference Number, pp. R169-R180 (November 30, 1965).
38. Barnhart, E., Application of Aerobic Digestion to Industrial Waste Treatment.
Proceedings-16th Industrial Waste Conference, Purdue University, Lafayette, Indiana,
pp. 612-618(1961).
39. Lawton, G.W., and Norman, J.D., Aerobic Digestion Studies. Journal Water Pollution
Control Federation, 36, No. 4, pp. 495-504 (1964).
40. Ritter, L., Design and Operating Experiences Using Diffused Aeration for Sludge
Digestion. Journal Water Pollution Control Federation, 42, No. 10, pp. 1,782-1,791
(1970).
41. Dreier, D.E., Discussion on Aerobic Sludge Digestion. By Jaworski, N., Lawton, G.W.,
and Rohlich, G.A. Presented at the Conference on Biological Waste Treatment,
Manhattan College, N. Y., N. Y. (April 20-22, 1960).
42. Levis, C.E., Miller, C.R., and Vosburg, L.E., Design and Operating Experiences Using
Turbine Dispersion for Aerobic Sludge Digestion. Journal Water Pollution Control
Federation, 43, No. 3, pp. 417-421 (1971).
43. Howe, R.H.L., What to do with Supernatant. Waste Engineering 30, No. 1, page 12
(1959).
44. Randall, C.N., and Koch, C.T., Dewatering Characteristics of Aerobically Digested
Sludge. Journal Water Pollution Control Federation, Research Supplemental, No. 5,
Part 2, pp. R215-R238 (1969).
45. Private communication with C.L. Swanson, U. S. EPA, Cincinnati, Ohio (November 12,
1970).
46. Personal communication with Kai W. Young, Union Carbide Corporation, Linde
Division, Tonawanda, N. Y. (March 29, 1974).
47. Kampelmacher, E.H., and van Noorle Jansen, L.M., Reduction of Bacteria in Sludge
Treatment. Journal Water Pollution Control Federation, 44, No. 2, pp. 309-313
(1972).
11-31
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48. Evans, S.C., Sludge Treatment at Luton. Journal Institute Sewage Purification, Part 5,
pp. 381-390 (1960).
49. Doyle, C.B., Effectiveness of High pH for Destruction of Pathogens in Raw Filter Cake.
Journal Water Pollution Control Federation, 39, No. 8, page 1,403 (1967).
50. Farrell, J.B., Smith, J.E., Jr., Hathaway, S.W., and Dean, R.B., Lime Stabilization of
ChemicalPrimary Sludge at 1.15 Mgd. Presented at the 45th Annual Conference of
the Water Pollution Control Federation, Atlanta, Georgia (October, 1972).
11-32
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CHAPTER 12
SLUDGE DEWATERING
12.1 General
The selection of a sludge dewatering process is mainly dependent on the subsequent
treatment and disposal of the dewatered sludge. If incineration is used, sludges with low
moisture content are required to minimize fuel consumption. However, if dewatered sludge
is taken directly to a landfill, higher sludge moisture contents may be tolerable depending
on the construction and management practices of the landfill operation. Other factors that
should be considered are space limitations, economics and general operation.
12.2 Vacuum Filtration
Vacuum filtration is the most common mechanical method of sludge dewatering utilized in
the United States. The majority of vacuum filters are installed in larger facilities where
sludge drying beds are impractical due to limited space, or where incineration is necessary
for maximum volume reduction. In small plants, vacuum filters have not been commonly
used because of the higher operating skill required; the increased construction costs for
facilities to house chemical storage, conditioning and filtration facilities; and the higher
operating cost for chemical conditioning. Today, however, there is an increasing trend
toward the use of vacuum filters in smaller communities, because of improvements in
operator training, the increasing scarcity of land and, in some instances, the difficulty of
obtaining unskilled labor for cleaning of sludge drying beds. A flow diagram of a typical
vacuum filter installation is shown on Figure 12-1.
12.2.1 Process Considerations
Process operating considerations for vacuum filtration include:
1. Type of sludge
2. Feed solids concentration
3. Sludge age
4. Chemical requirements for sludge conditioning
5. Sludge mixing and flocculation
6. Drum speed and drum submergence
7. Filter fabric characteristics
8. Filter cake discharge.
12-1
-------
FIGURE 12-1
TYPICAL VACUUM FILTER FLOW DIAGRAM
to
COAGULANT FLOW CONTROL
SLUDGE
C
o
)
_!
POLYELETROLTYE
. 9l C 1
=3
/
A
CJ
=>
s~
*
«
0
" \.
*
SLUDGE CONDITIONING TANKS
-DRUM
FILTRATE RETURNED
TO PLANT
AIR TO
ATMOSPHERE
SILENCER
WATER TO PLANT
CONVEYOR / | <VACUUM RUMP
FILTRATE/ WATER
PUMP
WASHINGS
RETURNED
TO PLANT
TO FILTER
CLOTH WASH
FROM WATER SOURCE
-------
Each of these parameters affects the filter yield, economy of operation and filter cake
characteristics. The nature of the sludge to be filtered is important in establishing several of
the design parameters. Sludge particle size, shape, consistency and density affect filterability
and chemical conditioning requirements (1). Because digestion results in the breakdown of
coarse solids into finer particles, vacuum filtration of raw primary or primary-secondary
sludges permits higher filter yields, lower chemical requirements and lower cake moisture
contents than vacuum filtration of digested sludges. As the ratio of secondary to primary
sludge increases, vacuum filtration of undigested sludge becomes more difficult.
The feed solids concentration has an important influence on the filtration rate and filter
yield. In a large number of cases where poor filter performance has been experienced,
sufficient care in operation has not been given to optimizing the feed solids concentration
(1). The optimum solids concentration for filtration is 8 to 10 percent. Higher solids content
makes the sludge difficult to pump and condition for dewatering. Lower solids content
requires the use of more or larger vacuum filters than necessary. In general, sludge yield
increases directly in proportion to an increase in feed solids concentration, although data
have been reported where the increase was more than proportional (2)(3)(4).
It is generally agreed that as raw sludge ages, it becomes more difficult to dewater on
vacuum filters. This effect has been reported by Ettelt and Kennedy (5), and Tenney, et al,
(3) who observed increasing filter cake moisture with increasing age of thickened activated
sludge. Tests at the Chicago Sanitary District (5) showed that even a short holding period
after thickening can significantly affect vacuum filter performance. It was observed that the
filter cake solids concentration varied linearly with detention time, decreasing from
17 percent for immediate filtering after thickening to 13 percent after a holding period of
3-1/2 hours. In this situation, it was found that freshening of the sludge by reaeration prior
to filtration dramatically reduced the filter cake moisture and the required ferric chloride
conditioning dosage.
Few, if any, raw or digested wastewater sludges can be successfully dewatered without some
form of chemical conditioning using ferric chloride, lime and/or polyelectrolytes. Proper
sludge conditioning requires a determination of optimum chemical dosages. Experience and
careful laboratory monitoring of the sludge characteristics are key factors in maintaining the
proper chemical proportions and concentrations. Where a mixture of undigested primary
and secondary sludge is to be dewatered, it is essential that careful attention be given to
maintaining a uniform ratio of primary to secondary sludge. Mechanically mixed and/or
aerated blending tanks are often provided for this purpose.
Optimum sludge mixing and flocculation under varying conditions require that sludge
conditioning tanks be provided with variable-speed mixer drives, adjustable weirs or flow
control devices to vary the sludge detention time and multiple points of chemical
application as indicated on Figure 12-1. The sludge slurry must be agitated sufficiently to
12-3
-------
maintain uniformity; however, excessive agitation should be avoided so that the conditioned
slurry particles are not sheared.
Drum speed and drum submergence are important factors in the operation of vacuum filters
since they affect filter yield and filter cake moisture. Increasing the drum submergence
generally results in increased filter yield, but produces a filter cake with higher moisture
content. Decreasing the drum speed, i.e., increasing the cycle time, has the opposite effect
of decreasing the filter yield, but produces a cake with lower moisture content.
Much information is available from the various manufacturers of vacuum filters and filter
media on the selection of proper media. A wide variety of media are now available including
natural and synthetic fabrics and metal coils. Because of the variable nature of sludge, it is
important that the selection of the media be based as nearly as possible on actual plant
conditions. Often this can best be accomplished through laboratory experimentation, using
the Filter Leaf Test, to provide information on filter fabric blinding, filtrate quality and
cake characteristics.
In general, a media is selected on the basis of the cleanest possible filtrate consistent with
high filtration rates, and where incineration is used, lowest moisture contents. In
incineration situations, it is often economically justifiable to oversize the filters to achieve
lower moisture contents and significant fuel savings. The concentration of total solids in the
vacuum filter filtrate is the primary measure of filtrate quality. The SS may vary between
100 and 20,000 mg/1 depending on sludge type, the filter media, and the applied vacuum
(1). In practice, filtrate SS concentrations of 500 to 2,000 mg/1 are normal. Although the
solids in the filtrate normally resettle readily when returned to the head of the plant, fine
solids may resist removal and be continuously recycled. Accumulation of these solids in the
system may eventually reduce overall plant efficiency. Activated sludges normally contain
more fine particles than primary sludge and, therefore, require tighter filter medias and
slower filtration rates.
A prerequisite for good filter operation is that the sludge cake discharge cleanly from the
filter media (6). Characteristics of both the sludge and the media affect filter cake discharge.
Primary sludges having high fiber content generally produce a drier cake that discharges
more readily than activated or digested sludges containing finer material. Because the cake
discharge characteristics of various fabrics differ, a number of materials should be evaluated
when running Filter Leaf Tests to determine the best fabric for a given sludge. Generally,
cake discharge occurs through the flexing action of the media passing around a
small-diameter discharge roller. Where sludge characteristics are such that the cake tends to
adhere to the media, a scraper blade may be provided to assist removal.
12-4
-------
12.2.2 Evaluation of the Vacuum Filtration Process
Experience has shown that there are considerable variations in filtration rate, not only
between different sludge types, but also between the same types of sludges at different
times and locations. The discrepancies in filter test results at different plants are usually
related to variations in feed solids concentrations, particle size distributions, and industrial
waste components in the raw wastewater. Obviously then, it is of major concern to have a
laboratory technique which can accurately predict the performance of a full-scale vacuum
filter prior to its installation.
The Buchner Funnel Test and the Filter Leaf Test are commonly used in laboratory testing
programs for determining the filterability of sludges. When the amount of representative
sludge is limited (less than 10 liters), it is advisable first to perform the Buchner Funnel Test
to determine optimal chemical dosage and sludge filtration characteristics. The Filter Leaf
Test can then be run at the optimum condition to determine filter yield. If a large amount
of sludge is available, the Buchner Funnel Test can be eliminated and the Filter Leaf Test
run instead.
The main purpose of the Buchner Funnel Test is to determine optimum chemical
requirements based on a comparison of the specific resistance of chemically treated sludge
with that of untreated sludge. An approximate filter yield can also be calculated from the
Buchner Funnel Test. Basically, the Buchner Funnel Test consists of filtering 100 ml of
sludge, either raw or conditioned, through filter paper under a vacuum of 20 to 25 inches of
mercury. The volume of filtrate (V, in ml) with time is noted and plotted against elapsed
time/volume (t/V in sec/ml) to obtain the slope of the resulting line. Using the above
information, the specific resistance of the sludge is calculated from the following equation:
r = 6.91 x 106
bA2P
where:
r = Specific resistance, sec2/g
b = Slope of plot (V vs. t/V), sec/ml2
A = Area of filter, sq cm
P = Filtration vacuum, psig
^ = Absolute viscosity of filtrate, centipoise
c = Initial SS concentration, mg/ml
The dimensions of the variables in the above equation are in units typically measured in the
laboratory. The conversion constant, 6.91 x 10^ reduces the variables to units which are
dimensionally consistent. The specific resistance as calculated by the above equation would
be expressed as sec2/g. Typical values of specific resistance for various sludges are shown in
Table 12-1.
12-5
-------
TABLE 12-1
TYPICAL SPECIFIC RESISTANCE VALUES FOR VARIOUS SLUDGES
Raw Sludge
5% FeClg
5% FeCl3 + 5% CaO
2% Polyelectrolyte
Digested Sludge
5% FeCl3
5% FeClg + 5% CaO
2% Polyelectrolyte
Primary Sludge
Specific Resistance,
107 sec2/g
1,300-2,100
10
3
3
400-1,600
25
18
20
Activated Sludge
Specific Resistance,
107 sec2/g
2,800
140
60
50
800
Reference
8
8
8
7
8
8
8
The approximate filter yield can be estimated from the specific resistance data using the
following relationship (9):
i (loo-CiA %
L = 0.0357
where:
L = Filter yield, psf/hr
Cj and Cf = Initial and final moisture content of the sludge, percent
m = Percentage of time for which vacuum acts during cycle
9 - Time for one drum revolution, minutes (usually between 1.5 and 5 minutes)
Y' - Absolute viscosity of filtrate, centipoise
R = r x lO-7 g/sec2
r = Specific resistance, sec2/g
P = Filtration vacuum, psig
The Filter Leaf Test techniques are simple, and the test can be run with minimum effort.
The advantage of the Filter Leaf Test is that the filter yield is measured and not merely
calculated using an empirical equation. With careful laboratory techniques, results will be
closely indicative of full-scale vacuum filter operation, although plant scale tests are more
conclusive and should be used whenever possible. A scale-up design factor of 0.9 is typically
used for the Filter Leaf Test where plant scale tests are not practical.
12-6
-------
The Filter Leaf Test is usually performed on a 0.1 sq ft filter leaf. The main objective of the
test is to evaluate the effect of different fabrics, fabric forms and drying times on filter
yield. Varying doses and types of chemicals should also be tested to establish chemical
conditioning requirements. The basic steps in performing a Filter Leaf Test are outlined
below and should be repeated a number of times to optimize the performance of each media
tested.
1. Condition the sludge in a 2 to 5 gallon container.
2. Submerge the filter leaf in the sludge slurry and apply vacuum for a fixed form
time.
3. Gently remove the filter leaf from the sludge slurry to allow the cake to dry for a
fixed drying time.
4. Remove the cake from the filter leaf and measure the weight and moisture
content of the cake. Note cake thickness and separability from the media.
5. Measure the filtrate SS and filtrate volume.
The filter yield is then calculated using the following equation:
L = dry sludge weight (g) x number of filtration cycles/hr
453.6 x area of test filter leaf (sq ft)
where:
L = filter yield in psf/hr
12.2.3 Vacuum Filter Performance
The results of studies by the Dow Chemical Company (10) on vacuum filter performance
are shown in Table 12-2. Various types of municipal sludges were dewatered using
polyelectrolytes and inorganic chemicals as conditioners for the same sludges. The data
shown in Table 12-2 indicate that the yield obtained when using polyelectrolytes for sludge
conditioning is generally higher than that obtained when using inorganic chemical
conditioners.
Vacuum filtration of pure oxygen waste activated sludge has been studied at Batavia, New
York, and in several other demonstration plants (11) (12). Data from these studies are
shown in Table 12-3. The settling characteristics of pure oxygen waste activated sludge are
such that, where primary settling is omitted, it can be thickened adequately in the clarifiers
for direct vacuum filtration. In this instance, additional sludge storage capacity beyond that
12-7
-------
TABLE 12-2
VACUUM FILTRATION PERFORMANCE USING INORGANIC CHEMICALS
AND PURIFLOC C-31 ON MUNICIPAL SLUDGE (10)
Type of Sludge
la Raw primary
b Raw primary
2a Raw primary
b Raw primary
3a Raw primary
b Raw primary
4a Raw primary
b Raw primary
5a Digested primary
b Digested primary
6a Digested primary
b Digested primary
7a Elutriated/digested/primary
b Elutriated/digested/primary
8a Elutriated/digested/primary
b Elutriated/digested/primary
9a Elutriated/digested/primary
b Elutriated/digested/primary
lOa Elutriated/chgested/primary
and secondary
b Elutriated/digested/primary
and secondary
lla Digested primary and
secondary
b Digested primary and
secondary
12a Elutriated/digested/primary
and secondary
b Elutriated/digested/primary
and secondary
Filter Media
Coil
Coil
Coil
Coil
Open synthetic
Open synthetic
Open synthetic
Open synthetic
Long napped dacron
Long napped dacron
44 x 44 Saran
44 x 44 Saran
Napped polyester
Napped polyester
Napped polyester
Napped polyester
Napped polyester
Napped polyester
Long napped dacron
Long napped dacron
Synthetic
Synthetic
Napped dacron
Napped dacron
Chemical
Conditioning
FeCl3
Lime
C-31
Fe2(S04)3
Lime
C-31
FeCl3
Lime
C-31
FeCl3
Lime
C-31
Fe2(S04)3
C-31
FeClg
Lime
C-31
FeCl3
C-31
FeCl3
C-31
Fe2(S04)3
C-31
Lime and
F"2(S04)3
C-31
FeClg
Lime
C-31
Fe2(S04)3
Lime
C-31
Solids Concentration
Dosage
Ib/ton dry solids
162.4
166.8
14.0
60
106
8.4
80.0
280.0
18.0
78.0
390.0
20.0
(16.06/T)
(S6.58/T)
66.0
206.0
17.0
56.8
10.2
100.0
8.0
108.0
9.0
(S8.90/T)
(S8.79/T)
610.0 (total)
or($10.07/ton)
18.0 or
($6.12/ton)
360.0
120.0
22.0
Initial
percent
7.57
7.0
9.1
9.6
15
12
11.2
10.7
7.2
7.2
15.0
15.0
6.1
7.7
10.4
10.1
10.9
11.1
9.1
9.1
4.4
4.3
8.0
9.0
Final
percent
20.1
20.0
24.0
20.0
40.0
30.0
39.0
34.5
26.0
33.5
43.0
32.0
36.0
32.9
32.7
38.6
34.0
35.0
25.0
24.0
26.2
23.7
28.0
25.0
Filter Yield
psf/hr
6.91
7.53
8.6
11.5
5.0
7.0
3.1
3.5l
3.9
11.0
9.2
25.0
5.74
12.66
3.88
5.94
2.8
5.5
4.7
7.4
5.2
5.6
5.1
7.25
1. Yield intentionally kept down to avoid overloading incinerator.
12-8
-------
Location
Batavia, N. Y.2
Louisville, Ky.
New Orleans, La.3
Philadelphia, Pa.3
Notes:
TABLE 12-3
SUMMARY DATA ON VACUUM FILTRATION OF
PURE OXYGEN AERATION SLUDGES (11)
Type
of
Sludge
Waste activated
Waste activated
Waste activated
Thickened waste activated
Waste activated
Waste activated
Combined 1
Waste activated
Waste activated
Waste activated
Waste activated
Feed Solids
percent
2.75
2.75
2.29
4.37
2.91
3.58
6.39
3.00
3.00
1.48
1.84
174
174
192
200
115
190
88
100
150
144
146
Chemical Dose Rate
Ib/ton of Dry Solids
FeCl3 Lime
272
266
1. Combined Sludge: 5 parts raw primary sludge, 1 part waste activated sludge.
2. Pilot Scale Filter (3 ft diameter by 1 ft length).
3. Filter Leaf (filter area = 0.1 sq ft).
Cake Yield
psf/hr
1.96
2.14
4.55
5.12
2.36
1.95
4.08
5.60
5.30
1.72
1.56
Cake
Solids
percent
24.1
20.5
16.0
14.5
13.3
17.6
29.0
16.3
19.5
30.8
36.0
-------
available in the clarifiers is necessary to allow optimum filter operation without adversely
affecting clarifier performance. The most effective chemical conditioner for vacuum filtering
pure oxygen sludge at Batavia was found to be ferric chloride at a dosage of 175 to
200 Ib/ton of dry solids (11). Cake yield was roughly doubled by increasing the drum speed
from 6.3 minutes per revolution to 2.5 minutes per revolution, but cake moisture content
increased substantially. The use of polyelectrolytes alone or in combination with inorganic
chemicals was not as effective as ferric chloride.
12.2.4 Upgrading Existing Vacuum Filters
In some instances, recycling a load of fine solids within the plant can seriously impair
vacuum filter performance. Such conditions are created by poor quality digester supernatant
or elutriate which must be corrected before good filter performance can be reestablished.
The need to upgrade existing vacuum filters is usually due to an increase in sludge
production or a change in sludge characteristics. In these cases, existing vacuum filter
operations may be upgraded by improving one of the following:
1. Feed solids concentration
2. Filter media
3. Sludge conditioning.
Sludge thickening to increase feed solids concentration to the vacuum filter is generally the
most economical and therefore the most desirable method of increasing the filtration rate.
An increase in feed solids concentration will normally result in at least a proportional
increase in the filtration rate (1) (4). Therefore, the effect of upgrading sludge thickening
facilities to increase feed solids concentration from 4 percent to 5 percent would be to
increase the filtration rate or sludge dewatering capacity by 25 percent. For this reason,
steps should be taken to ensure that feed solids concentration is maximized before other
upgrading techniques are considered.
The use of proper vacuum filter media is very important to efficient filter performance.
Most primary sludges and certain industrial waste sludges have fibrous, non-uniform solids
that may clog or blind improper media and lead to lower filter yields, increased chemical
consumption and the need for frequent washing (1). For this reason, primary sludges are
most effectively dewatered with media having comparatively large openings that resist
blinding. These media, however, are not suitable for activated or digested sludge due to the
higher proportions of fine particles. These sludges require tighter media and lower filtration
rates. Therefore, where sludge characteristics have been altered significantly due to
upgrading procedures, vacuum filter operation often may be correspondingly upgraded by
replacing filter media as indicated by laboratory Filter Leaf Tests or full-scale trials (1).
12-10
-------
Chemical conditioning is normally required prior to vacuum filtration of sludge. The
operation of a dewatering facility, therefore, may be upgraded by providing more efficient
chemical sludge conditioning. Most older filter installations were designed to use inorganic
chemicals, such as ferric chloride and lime, as sludge conditioners. Within the last 10 years,
organic polyelectrolytes have begun to replace inorganic chemicals as sludge conditioners,
and offer an attractive advantage in more economical storage, handling and feeding
equipment. Polyelectrolytes are also less corrosive and frequently less expensive than
inorganic chemicals. The use of polyelectrolytes in improving the operation of vacuum
filters has been practiced at treatment plants in Bay City, Michigan (13), and Kansas City,
Missouri (14), among others.
The Bay City Wastewater Treatment Plant provides primary treatment for 7.0 mgd and
produces approximately 450 tons of dry solids/year. Until 1961, raw primary sludge was
conditioned with ferric chloride and either kiln-dried pebble lime or spent carbide (calcium
hydroxide formed as the result of chemical action in making acetylene). Bay City's
conditioned sludge is dewatered on vacuum filters having an effective area of 150 sq ft. The
vacuum filter cake is incinerated. In 1961, polyelectrolytes were tried as sludge conditioners
in an attempt to improve filter yield. Results of filter operation for 1959-1964 using ferric
chloride, lime and polyelectrolytes are presented in Table 12-4. These results clearly indicate
that use of polyelectrolytes increased the filter yield and significantly reduced vacuum
filtration operation time. The cost of chemicals for sludge conditioning at Bay City was
$10.40/ton of dry solids when using ferric chloride and kiln-dried lime, $7.15/ton of dry
solids when using ferric chloride and carbide lime and $7.30/ton of dry solids when using
polyelectrolytes. The following advantages were realized when polyelectrolytes were used
for sludge conditioning:
1. Savings in equipment and floor space
2. Improved housekeeping
3. Improved safety
4. Reduced quantities of ash, with a large reduction in ash handling and storage
5. Reduction in operating time, with resulting savings in operating and maintenance
costs.
12.2.5 Process Design and Cost Estimates
The following example will serve to illustrate design and cost considerations when upgrading
vacuum filter installations by converting from inorganic chemical to polyelectrolyte sludge
conditioning.
12-11
-------
TABLE 12-4
A COMPARISON BETWEEN LIME/FERRIC CHLORIDE AND POLYELECTROLYTES
FOR CONDITIONING RAW PRIMARY SLUDGE (13)
to
Year
1959
1960
1961-62
1962-63
inAQ A/I
Dry
Solids
tons
461
580
424
415
/I 37
Feed
Solids
percent
11.2
11.2
10.9
10.7
in o
Filter Chemical Added, Ib
Yield Lime FeClo. C-31 C-32 C-149 A-21
psf/hr
3.1 162,000 31,000 - -
3.1 225,000 44,000 - -
5.3 - 5,562
5.5 - - 10,300 -
Total of all three
polyelectrolytes
A3 U- 7A71 «J
Cake
Solids
percent
40.1
39.0
35.9
34.5
34 fi
Solids
Recovery
percent
64.1
62.1
75.6
73.7
75 Q
Operation
Time
hr/yr
2,125
2,420
1,119
1,114
1 .301
-------
An existing vacuum filter installation was annually conditioning and filtering 300 tons (dry
basis) of mixed digested primary and secondary sludge with a filter yield of 3 psf/hr. Sludge
conditioning had been accomplished using 65 Ib of ferric chloride and 200 Ib of lime/ton of
dry solids. To reduce the costs involved in bulk chemical handling and to increase filter
yield, the applicability of a polyelectrolyte system for sludge conditioning was investigated.
The optimum polyelectrolyte dosage was found to be 20 Ib/ton of dry solids, added to the
digested sludge in a 1-percent solution. The polyelectrolyte addition resulted in a
subsequent filter yield of 4 psf/hr. This upgrading procedure reduced the operating time of
the vacuum filter by 25 percent, thereby decreasing operational and maintenance costs.
The capital cost for the polyelectrolyte application system was estimated at $7,000. This
cost included all required tankage, pumps, and mixers.
12.3 Drying Beds
The dewatering of digested sludge on drying beds has long been practiced in the United
States. Historically, sludge drying beds have mainly been used for smaller communities (15).
The widespread use of drying beds for smaller plants is due to their simplicity and low
maintenance costs. The chemical cost and operating complexity of mechanical dewatering
equipment are additional factors which favor drying beds. Disadvantages include their large
land requirement, inability to dewater effectively during inclement weather, difficulty of
obtaining labor for cleaning, and potential odor problems.
12.3.1 Process Considerations
One of the difficulties in developing a rational design for sludge beds is the multitude of
variables which affect the sludge drying rate. In practice, it is difficult to isolate these
variables and evaluate them quantitatively. Some of the more important variables are (16):
1. Climate and atmospheric conditions
a. Temperature
b. Humidity
c. Rainfall
d. Wind velocity
e. Barometric pressure
f. Solar radiation
12-13
-------
2. Sludge characteristics
a. Type of sludge
b. Method of stabilization
c. Moisture content
d. Grease content
3. Operational factors
a. Sludge age
b. Presence or absence of coagulants
c. Depth and frequency of application
4. Bed characteristics
a. Condition and gradation of sand
b. Condition of drainage system
c. Travel distance to bed extremities.
Notwithstanding the magnitude of the problems involved, some generalizations concerning
the applicability of these factors can be made. When possible, decisions regarding the
specific effect of any or all of these factors should be based on bench-scale testing.
Quon and Johnson (17) have indicated that well-digested sludge should be applied to drying
beds in depths of 6 to 9 inches, with 8 inches appearing to give optimum drying rates.
Sludge should be properly digested before being applied to the drying beds. Raw or poorly
digested sludge dewaters slowly and produces strong odors. Sludge that has been overly
digested will contain a high percentage of fine solids which will impair drainage. Aerobically
digested sludge usually drains well on sand drying beds (1) (18).
Drainage and evaporation are the mechanisms which affect the dewatering rates of
well-digested sludge on drying beds. It has been widely accepted that under normal
conditions, practically all of the drainage of sludge occurs during the first three days
following the filling of the drying bed (6). After this initial period, it was felt that
evaporation was largely responsible for additional dewatering of the sludge. Recent studies
indicate this is not the case (16) (17) (19). In an extended study, it was found that the
initial rate of drainage was small, but that it increased with time (17). After approximately
three days, the drainage rate increased and the sludge surface dropped substantially. This
phenomenon is explained by considering that air trapped in the voids of the sand bed is not
12-14
-------
free to move and thus impedes the initial flow of water through the filter. Eventually, this
air is liberated, allowing a greater flow to pass through the sand bed. After a period of
maximum drainage, the drainage rate gradually decreases due to the buildup of solids on the
sand surface, which offers resistance to further filtration. Once this point is reached,
evaporation from the free water surface accounts for further dewatering.
Drying beds are usually constructed with an underdrain piping system to increase the
drainage rate. Although some installations have been designed with asphalt or concrete
bottoms to facilitate sludge removal, experiments have shown that drained sludge beds dry
25 percent faster than beds with an impervious bottom (17). Underdrainage from drying
beds should be returned to the plant for treatment (20).
In areas with adverse climatic conditions, the use of glass-covered beds, while expensive, has
been found to increase the total output of dewatered sludge by 100 percent (21). Recent
work in northern Texas, however, indicated that during the dry season, covers retarded the
drying rate rather than accelerating it (16). Thus, where covered beds are constructed,
adequate ventilation must be provided so that maximum evaporation rates may be
maintained.
12.3.2 Design Basis
Present-day design practices are based largely on comparisons with existing plants in the
area, or upon empirical recommendations. The sludge drying bed area requirements shown
in Table 12-5 are recommended for domestic wastewater treatment plants located in the
northern United States (6).
TABLE 12-5
SLUDGE-DRYING BED AREA REQUIREMENTS
Type of Sludge
Primary Digested
Primary and Humus Digested
Primary and Activated Digested
Primary and Chemically Precipitated Digested
Area of Drying Beds
sq ft/capita
Open Beds
1.0 to 1.5
1.25 to 1.75
1.75 to 2.5
2.0 to 2.5
Covered Beds
0.75 to 1.0
1.0 to 1.25
1.25 to 1.5
1.25 to 1.5
12-15
-------
In the southern United States, reduced areas are often practical because of more favorable
climatic conditions.
Recently, rational design procedures have been developed for sludge drying beds (6).
Equations which describe a sludge's drainage and drying properties have been derived (22).
These equations lend themselves to incorporation in computer models for sludge properties,
weather variability and economics. These methods may eventually replace the present
empirical practice for designing drying beds.
12.3.3 Upgrading Existing Facilities
It is possible to upgrade an overloaded sludge drying bed by the following methods:
1. Improving the performance of upstream facilities, e.g., thickeners, digesters
2. Adding chemicals to improve sludge dewatering characteristics
3. Covering open beds where climatic conditions adversely affect performance.
Chemicals such as alum, ferric chloride and polyelectrolytes have been used as flocculants to
improve the dewatering capacity of sludge drying beds. The use of these chemicals increases
the permissible sludge loadings to the drying beds by increasing the number of bed
applications per year.
In general, the chemical flocculants allow greater amounts of water to drain from the sludge,
thereby decreasing the amount of water to be removed through the slower evaporation
process. Bed loadings for chemically treated and untreated sludge should be evaluated by
laboratory and field testing to determine the effectiveness of chemical addition on sludge
drying. Buchner Funnel Tests can be used to predict dewatering rates on drying beds in the
same manner as for vacuum filters (23). Care must be taken to avoid adding excess amounts
of chemicals, which might blind sand particles and lower drainage rates. Alum has been used
successfully at a dosage of 1 lb/100 gal of digested sludge (1). On the other hand,
polyelectrolyte has been used at dosages as low as 0.05 lb/100 gal of digested sludge (16).
12.3.4 Process Designs and Cost Estimates
An open drying bed at an existing activated sludge plant was originally designed based on a
population equivalent of 20,000 and a bed area requirement of 2 sq ft/capita and was
loaded at a rate of 10 Ib of dry solids/sq ft/yr. As a result of upgrading secondary treatment
units, it was necessary to increase the loading to 15 Ib/sq ft/yr to accommodate the
increased sludge quantities. Two alternatives were available for upgrading the existing drying
beds. It was possible either to cover the beds or to add 1 Ib of alum/100 gal of digested
sludge to decrease the drying time by approximately 50 percent.
12-16
-------
Covering the drying beds was estimated to cost $240,000. The estimated cost of the alum
slurry feed system and flocculation tank was $33,000.
Based on comparison of these capital costs, it appears that chemical addition may be the
most economic alternative. However, to draw a definite conclusion, it would be necessary to
compare total annual costs of both alternatives over the expected life of the facility.
12.4 Centrifugation
The use of centrifuges for separating materials of different densities is well established as
discussed in Section 10.5. However, their use in the United States for sludge dewatering is
not as widespread as the use of vacuum filters. Recent improvements in centrifuge design
and the availability of full-scale performance data have encouraged their use for sludge
dewatering.
12.4.1 Types of Centrifuges
There are three general classifications of centrifuges that can be applied to sludge
dewatering: solid-bowl, disc, and basket. These are illustrated on Figure 12-2. Typical
centrifuge characteristics are summarized in Table 12-6.
TABLE 12-6
SUMMARY OF CENTRIFUGE CHARACTERISTICS (25)
Centrifuge Description
Description
Method of Operation
Bowl Diameter, in.
Flow Rate, gpm
Solids in Feed, percent
Solids Discharged
Speed, rpm
Centrifugal Force, G
Motor Horsepower
Solid-Bowl
Continuous
6 to 60
1 to 200
*
1 to 15 tons/hr
1,000 to 6, 000
3,200 max
5 to 250
Basket
Batch
12 to 60
100 max
0.1 to 30
l,0001b-max
2,500 max
2,000 max
100 max
Disc
Continuous
8 to 30
10 to 300
0.1 to 10
10 to 3,000 gal/hr
4,500 to 10,000
12,000 max
10 to 125
*Any liquid or slurry which can be pumped.
12-17
-------
FIGURE 12-2
TYPES OF CENTIFUGES (24)
GEAR BOX
r
DRIVE SHEAVE
FEED
CENTRATE CAKE
DISCHARGE
SOLID-BOWL CENTRIFUGE
CAKE DISCHARGE «-
FEED
S3 it
CENTRATE
BASKET CENTRIFUGE
FEED
CENTRATE
UNDERFLOW
DISC CENTRIFUGE
12-18
-------
Solid-bowl centrifuges are widely used because of their dependable performance and
relatively low maintenance requirements. Solid-bowl centrifuges are suited to dewatering
primary sludge and mixtures of primary and waste biological sludge. For most sludges, to
achieve solids recovery in the range of 80 to 95 percent with a solid-bowl centrifuge requires
the addition of polyelectrolytes.
Unlike solid-bowl centrifuges, basket centrifuges operate on a batch basis. Because of the
large bowl diameter, basket centrifuges operate at slower speeds. The centrifuge can be
operated on an automatic cycle for programmed filling and skimming. Basket centrifuges are
generally applicable only for sludge dewatering at smaller plants.
Disc centrifuges are used primarily for thickening of waste activated sludge as discussed in
Chapter 10.
12.4.2 Process Considerations
Major process variables for centrifugation are feed rate, sludge solids characteristics, feed
consistency, temperature and chemical additives. Solid-bowl machine variables are bowl
design, bowl speed, pool volume and conveyor speed (1). Major performance factors are
cake dryness and solids recovery. To increase cake dryness, the following actions should be
considered (1):
1. Increase feed rate.
2. Decrease feed solids concentration.
3. Increase temperature.
4. Do not use flocculants.
5. Increase bowl speed.
6. Decrease pool volume.
7. Decrease differential conveyor speed.
Actions which should be considered for increasing solids recovery are as follows (1):
1. Decrease feed rate.
2. Increase feed solids concentration.
3. Increase temperature.
4. Use flocculants.
5. Increase bowl speed.
6. Increase pool volume.
7. Decrease differential conveyor speed.
12-19
-------
Most of the variables which improve cake dryness tend to decrease the solids recovery. This
is an important feature of centrifuge operation. Therefore, operation of a centrifuge should
be optimized to obtain the desired balance between cake dryness and solids recovery.
The following advantages are associated with solid-bowl centrifugal dewatering:
1. Capital cost is less than for vacuum or pressure filters.
2. The unit is totally enclosed so that odors are more readily controlled.
3. The unit will fit in a small space and requires a minimum of auxiliary equipment.
4. A wide variety of solids can be handled.
5. Minimum operator attention is required.
Disadvantages of solid-bowl centrifugal dewatering are:
1. Solids capture is often poor without the use of chemicals.
2. Chemical costs can be substantial.
3. Cake moisture is often higher and centrate quality lower than with vacuum
filtration.
4. Maintenance costs are high, especially if the sludge contains substantial grit
quantities.
5. Fine solids which escape in the centrate may resist settling when recycled to the
head of the treatment plant, gradually build up in concentration and eventually
raise effluent solids level. However, this should not be a problem if capture
efficiencies can be maintained above 90 percent.
12.4.3 Design Considerations
Centrifuges should be selected on the basis of pilo^ tests with smaller, geometrically-similar
machines. Proper scale-up for predictable results must consider and provide for the
following variables (26):
1. Physical nature of solids being handled
2. Stability of centrifuge feed
12-20
-------
3. Solids dewatering time
4. Chemical flocculant dosages
5. Percent solids recovery
6. Resistance to abrasion
7. Wet cake discharge rate.
The use of a small continuous pilot centrifuge for testing purposes may not be possible if
sludge supply is limited. Several manufacturers have correlated full-scale unit performance
with "spin tests" conducted on laboratory centrifuges. This technique has limited value in
estimating design criteria and generally is of use only to the manufacturer.
12.4.4 Centrifuge Performance in Sludge Dewatering
Table 12-7 contains operating data for solid-bowl centrifuges for various combinations of
municipal wastewater sludges.
Raw primary and digested primary sludges dewater easily; with polyelectrolyte addition, a
centrifuge can produce 25 to 40 percent cake solids with better than 90 percent recovery.
Waste activated sludge alone is very difficult to dewater. When waste activated or trickling
filter sludge is added to raw or digested primary sludges, the cake solids content is reduced
to about 20 to 25 percent, and the polyelectrolyte dosage to obtain 90 percent recovery
increases. Factors responsible for this loss in efficiency include lower feed solids
concentration and the poorer dewatering characteristics of biological sludge.
The use of centrifuges for sludge dewatering has been considered recently by several
municipalities. At one large southeastern city, a centrifuge test program was conducted to
determine the applicability of centrifugal dewatering of raw primary, digested primary,
co-settled raw primary and waste activated, mixed digested and primary digested plus
thickened waste activated sludges (27). With a solid-bowl centrifuge, 55 to 85 percent
recovery was obtained without the use of polyelectrolytes, depending upon the feed rate.
Recovery levels of 85 percent or better were achieved with 0.5 to 5.0 Ib of strong cationic
poly electrolyte/ton of dry solids for combined raw primary and waste activated sludge,
primary digested plus waste activated sludge and mixed digested sludge. Problems were
encountered in trying to use a solid-bowl centrifuge on the combination of thickened waste
activated and primary digested sludge. Higher levels of polyelectrolyte were required as the
proportion of waste activated sludge to primary sludge increased. When operating with a
mixture of two-thirds primary digested sludge and one-third thickened waste activated
sludge on a dry solids basis, 50 to 60 percent recovery was achieved without
polyelectrolytes. To increase recovery to 80 to 90 percent, polyelectrolyte dosages of 10 to
20 Ib/ton of dry solids were required.
12-21
-------
TABLE 12-7
SOLID-BOWL CENTRIFUGE PERFORMANCE DATA
Type of Sludge Capacity
gpm
Pnmary - Raw
23-26
23-36
27.5
275
27.5
Pnmary - Digested
70-140
Raw Pnmary and Trickling Filter
99-22
Digested Primary and Trickling Filter
Raw Pnmary and Waste Activated 40-80
30-80
Thickened Waste
Activated Sludse 40-60
(by Disc Onlnfup-) 30-80
digested Primary and V, astc \ctlvatcd 50-1 20
30-150
Digested Primary and Thn kcne d 40- 1 40
Waste Activated 30 80
Aerobic Digested
(Contact stabilization)
Heat Treatment Sludge
Zimpro
Porteut,
Chemical Sludges
Lime-Phosphate
Lime-Treatment in Primaries
Tertiary Phosphate Removal
(Chemical and Waste Activated)
Feed Cake
Solids Solids
percent percent
3040
30-40
25-35
28-40
30-40
9-12 28-50
912 28-50
3237
25-28
26-37
8.8
84
9.2
30-40
30-40
2430
25-35
20-30
6 22-28
6 22-28
26-30
22-30
4 > 17
4 > 17
24-30
25-35
96 23
2026
21-25
20-25
5 5 25+
5.6 20 2
2028
22-36
25
18-25
18-25
18-20
18-20
1523
19-25
18-24
18-24
5-6 15-17
5-6 15-17
22-30
2532
28 15-19
1520
15-20
18-24
1824
6 2033
6 20-22
1014
10 14
JO-45
3050
3540
40-45
1620
Recovery
percent
7080
95
90+
70-90
5090
65-80
80-98
84-93
88-95
85-96
44
84
97
70-90
95
90+
70-85
85-90
75-85
85
81-90
87-95
20-50
85
90+
65-75
96-100
82-96
83-90
85-94
40
90
90+
60-75
85
6085
95+
40-60
85
90+
5075
50-80
95
50-65
85
80-95
50-75
8590
50-60
85-100
50-70
95
40-80
85
90+
50-60
85-90
85-90
75-85
96-98
85-90
Polyclc
clrolyte
Requirement
t/ton
2-4
None
None
64
3-5
27
None
394
572
3-6
None
8
3-7
3-6
-
4-9
5-8
4-10
6-9
None
674
4-8
-
None
8-16
5 10
None
None
6-20
None
12
None
10-85
None
10-20
' None
13 17
5-10
None
None
None
None
Ib/ton
None
5-10
1 5-2 5
1.0-2 5
None
5-10
3-6
0
05-1
None
3-8
2 5-3.5
None
4-5
6-17
11-17
24-10
10
34
Reference
27
27
27
27
27
27
27
28
28
28
29
29
29
30
30
30
30
27
27
27
28
27
27
30
30
28
28
28
28
29
29
30
30
27
28
28
27
27
30
30
31
31
27
27
27
27
30
30
30
31
31
27
27
30
30
30
30
30
27
30
12-22
-------
At a southwestern community, field tests were conducted to determine the applicability of
centrifugation for dewatering combined primary and secondary digested sludge (27). The
results indicated that a solid-bowl centrifuge could be used to replace concrete drying beds.
With the use of a strong cationic polyelectrolyte at a concentration of 3 to 4 Ib/ton of dry
solids, the sludge could be dewatered to 17 to 18 percent solids with a solids recovery of
85 percent. Solids recovery was increased to 98 to 99 percent when the polyelectrolyte
dosage was raised to 5 to 6 Ib/ton of dry solids.
In El Paso, Texas, a land problem necessitated replacement of sludge drying beds (32).
Centrifuges were used to dewater the digested sludge at a 6 to 7 percent feed solids
concentration. The centrifuges produced 20 to 22 percent cake solids with 85 to 90 percent
recovery, at a polyelectrolyte dosage of 2 to 3 Ib/ton of dry solids.
The use of a cationic polyelectrolyte was evaluated for a combination of 80 percent raw
Imhoff and 20 percent digested sludge (27). The cationic polyelectrolyte was effective in
improving sludge dewatering at economical dosage levels (2 to 3 Ib/ton of dry solids). With
an average feed solids concentration of 8 percent, a cake solids of 35 percent at 95 percent
solids recovery was obtained.
12.4.5 Use of Centrifugal Dewatering for Upgrading Sludge Handling Facilities
The performance of centrifuges in various applications clearly indicates that centrifugation
should be considered when the upgrading of sludge dewatering facilities is required.
Centrifuges can be used to replace or supplement vacuum filters for sludge dewatering,
especially where space is limited. Where centrifuges are considered for an upgrading
situation, the efficiency of existing grit and screenings removal facilities must be carefully
studied. Solid-bowl centrifuge manufacturers normally recommend that cyclonic grit
removal be provided ahead of the centrifuge to reduce abrasion. Inadequate removal or
comminution of screenings can result in plugging of the centrifuge inlet. As centrifuges must
be periodically dismantled for maintenance, provision for an overhead hoist is desirable.
12.4.6 Process Design and Cost Estimates
The following example is presented to illustrate the upgrading of sludge handling facilities
through the use of centrifuges.
Vacuum filter facilities at an existing activated sludge plant were overloaded due to an
increase in plant flow from 10 to 20 mgd. Space limitations at the plant prohibited the
installation of additional vacuum filters.
It was decided to add centrifugal dewatering to handle the additional 12,000 Ib/day of dry
solids (20 gpm by volume). Facilities for sludge dewatering included two solid-bowl
centrifuges, each capable of handling 15 to 20 gpm, sludge feed pumps, polyelectrolyte
12-23
-------
addition facilities and other necessary appurtenances. The solid-bowl centrifuges were
designed to produce 20 to 25 percent cake solids at 85 percent recovery, with 5 percent
feed solids and a polyelectrolyte dosage of 3 to 6 Ib/ton of dry solids. The capital cost for
this upgrading procedure was estimated at $256,000.
12.5 Filter Presses
A variety of filter presses have been used to dewater sludge. The most commonly used type
consists of a series of rectangular plates, recessed on both sides, and supported face to face
in a vertical position on a frame with a fixed and movable head. A filter cloth is hung or
fitted over each plate. The plates are held together with sufficient force to withstand the
pressure applied during the filtration process. Hydraulic rams or powered screws are used to
hold the plates together.
In operation, chemically conditioned sludge is pumped into the space between the plates at
a pressure of 60 to 180 psig. Under this pressure, the liquid is forced through the filter cloth
and plate outlet ports. The plates are then separated and the sludge drops to a hopper
below. Filtrate normally is returned to the headworks of the treatment plant.,The sludge
cake thickness varies from 1 to 1-1/2 inches and the moisture content from 55 to
70 percent. A complete filtration cycle includes time required for pressing, cake removal,
media washing and press closing. Total filtration time varies from three to six hours with
pressing time ranging from one to three hours.
The major operating costs associated with this method of dewatering are those for chemical
conditioning and maintenance and replacement of filter cloths. Although not commonly
used in the United States, filter presses have been used extensively in Europe.
The advantages of filter presses are as follows:
1. Low cake moisture permits incineration of primary/secondary sludge
combinations without auxiliary fuel.
2. Continuous operating attendance is less than that required for vacuum filters.
3. A large filtration area can be installed in a minimum of floor area.
4. Hard to dewater sludges can be handled more readily.
Major disadvantages are as follows:
1. Capital costs are considerably higher than for centrifuges or vacuum filters.
2. Batch discharge requires equalization of pressed cake production prior to
incineration.
12-24
-------
12.6 References
1. Burd, R.S., A Study of Sludge Handling and Disposal. Federal Water Pollution Control
Administration, Publication WP-20-4 (May, 1968).
2. Sherwood, R.J., and Dahlstrom, D.A., Economic Costs of Dewatering Sludges by
Continuous Vacuum Filtration. Presented at the 65th Annual Meeting of the American
Institute of Chemical Engineers, New York, N. Y. (November, 1972).
3. Tenney, M.W., Echelberger, W.F., Jr., Coffey, J.J. and McAloon, T.J., Chemical
Conditioning of Biological Sludges for Vacuum Filtration. Journal Water Pollution
Control Federation, 42, No. 2, Part 2, pp. Rl-20 (1970).
4. Schepman, B.A., and Cornell, C.F., Fundamental Operating Variables in Sewage Sludge
Filtration. Sewage and Industrial Wastes, 28, 1443 (1956).
5. Ettelt, G.A., and Kennedy, J., Research and Operational Experience in Sludge
Dewatering at Chicago. Journal Water Pollution Control Federation, 38, No. 2, pp.
248-257(1966).
6. Sludge Dewatering. Water Pollution Control Federation Manual of Practice No. 20,
Washington, B.C. (1969).
7. Ford, D.L., General Sludge Characteristics. Included in Water Quality Improvement by
Physical Chemical Processes. Edited by E. Gloyna and W.W. Eckenfelder, Jr., Austin,
Texas University of Texas Press (1970).
8. Malina,, J. F., Sludge Filtration and Sludge Conditioning. Included in Water Quality
Improvement by Physical Chemical Processes. Edited by E. Gloyna and W.W.
Eckenfelder, Jr., Austin, Texas University of Texas Press (1970).
9. Eckenfelder, W.W., and O'Connor, D.J., Biological Waste Treatment. New York:
Pergamon Press (1961).
10. Sludge Conditioning with Purifloc. Dow Chemical Company (1966).
11. Young, K., Status of UNOX Sludge Pretreatment and Dewatering. Linde Division of
Union Carbide Corporation (September, 1971).
12. McDowell, M.A., Vahldieck, N.P., Wilcox, E.A., and Young, K.W., Continued
Evaluation of Oxygen Use in the Conventional Activated Sludge Process. U. S. EPA
Contract No. 14-12-867, Project No. 17050 DNW (February, 1972).
12-25
-------
13. Sherbick, J.M., Synthetic Organic Flocculants Used for Sludge Conditioning. Journal
Water Pollution Control Federation, 37, No. 8, pp. 1,180-1,183 (1965).
14. Hopkins, G., and Jackson, R., Polymers in the Filtration of Raw Sludge. Journal Water
Pollution Control Federation 43, No. 4, pp. 689-698 (1971).
15. Statistical Summary 1968 Inventory Municipal Waste Facilities in the United States.
Federal Water Quality Administration: Government Printing Office (1971).
16. Jennett, J.C., and Santry, I., Jr., Characteristics of Sludge Drying. Journal of the
Sanitary Engineering Division, ASCE, 95, No. 5, pp. 849-863 (1969).
17. Quon, J., and Johnson, G., Drainage Characteristics of Digested Sludge. Journal of the
Sanitary Engineering Division, ASCE, 92, No. 2, pp. 67-82 (1966).
18. Randell, C.W., and Koch, C.T., Dewatering Characteristics of Aerobically Digested
Sludge. Journal Water Pollution Control Federation, 41, No. 5, Part 2, pp. R215-238
(1969).
19. Jennett, J.C., and Harris, D.J., Environmental Effects on Sludge Drying Bed
Dewatering. Journal Water Pollution Control Federation, 45, No. 3, 449 (1973).
20. Recommended Standards for Sewage Works. Great Lakes-Upper Mississippi River
Board of State Sanitary Engineers (1971).
21. Nebiker, J., Drying of Wastewater Sludge in the Open Air. Journal Water Pollution
Control Federation, 39, No. 4, pp. 608-626 (1967).
22. Adrian, D.D., Dewatering Sludge on Sand Beds. Presented at the 65th Annual Meeting
of the American Institute of Chemical Engineers, New York, N. Y. (November, 1972).
23. Nebiker, J.H., Sanders, T.G., and Adrian, D.D., An Investigation of Sludge Dewatering
Rates. Journal Water Pollution Control Federation, 41, No. 8, Part 2, pp. R255-R266
(1969).
24. Lawson, George, R., Equipment and Chemicals - An Approach to Water Pollution.
Investment Dealer's Digest (August 5, 1969).
25. Townsend, Joseph, What the Was'ewater Plant Engineer Should Know about
Centrifuges. Water and Wastes Engineering, 6, No. 11, pp. 41-44 (1969).
26. White, W.F., Fifteen Years of Experience Dewatering Municipal Wastes with
Continuous Centrifuges. Bird Machine Company, Inc.
12-26
-------
17. Private communication with George Patenaude, Philadelphia District Representative,
Sharpies-Stokes Division, Pennwalt Corporation, Wynnewood, Pennsylvania (October
27, 1970).
28. Albertson, 0., and Guidi, E., Centrifugation of Waste Sludges. Journal Water Pollution
Control Federation, 41, No. 4, pp. 607-628 (1969).
29. Hercofloc Flocculant Polymers for Use in Sludge Conditioning. Hercules Incorporated,
Environmental Services Division, Bulletin ESD-102A, Wilmington, Delaware (1969).
30. Private communication with Gene Guidi, Sales Manager, Environmental Control
Equipment, Bird Machine Company, Walpole, Massachusetts (February 22, 1971).
31. Albertson, 0., and Guidi, E., Advances in the Centrifugal Dewatering of Sludges. Water
and Sewage Works, 114, No. 11, pp. 133-142 (1967).
32. El Paso Loses Drying Beds in Boundary Action. Water and Sewage Works, 117, No. 2,
pp. 26-27 (1970).
12-27
-------
-------
CHAPTER 13
CASE HISTORIES OF TREATMENT PLANT UPGRADING
13.1 Case History No. 1 - Use of Roughing Filter to Upgrade an Existing Low-Rate
Trickling Filter Plant (1) (2) (3)
Case History No. 1 involves the upgrading of an existing low-rate trickling filter plant in
Huber Heights, Ohio. The original plant was designed in August, 1956, for a flow of
0.7 mgd, with 85 percent BOD and SS removals. The community developed so rapidly that
by 1970, the average flow had increased to 2.3 mgd. The flow diagrams for the original and
upgraded plants are shown on Figure 13-1.
Operating and performance data for the overloaded plant for 1962, when the plant was
receiving 1.15 mgd, are compared with corresponding data after the plant was upgraded to
2.3 mgd in Table 13-1.
The comminutor and primary clarifiers were replaced with three Hydrasieve units of 1-mgd
capacity each. These units are stationary screens capable of removing 20 to 35 percent of
the BOD and SS in raw wastewater. A Hydrasieve unit is illustrated on Figure 13-2, along
with a flow diagram through the unit. These screens generally require no power and little
maintenance; however, a head loss does occur across the units.
The plastic-media roughing filter used in the upgrading operates at an application rate of
approximately 2.5 gpm/sq ft. Present BOD removal is about 25 to 35 percent across the
roughing unit. Because of the increased hydraulic loading, it was necessary to expand the
secondary clarification and chlorine contact tank capacities. The abandoned primary
clarifiers were converted to sludge thickeners. This step, in addition to conversion of the
low-rate anaerobic digester to a high-rate unit using gas recirculation for mixing, enabled the
sludge handling system to process the increased quantity of sludge produced.
This case history emphasizes that an existing plant may be gradually upgraded to handle a
three-fold increase in flow with the use of innovative techniques and newly applied process
equipment. The capital costs for upgrading the capacity of the plant were estimated at
approximately $352,000.
13.2 Case History No. 2 - Upgrading an Existing High-Rate Trickling Filter by Conversion
to a Super-Rate Filter System (4)
The North Treatment Plant at Sedalia, Missouri, a high-rate trickling filter plant designed for
1.25 mgd, was achieving 85 percent BOD removal in 1963. However, the State of Missouri
13-1
-------
FIGURE 13-1
CASE HISTORY NO.l
COMPARISON OF ORIGINAL AND UPGRADED FLOW DIAGRAMS
LEGEND
ORIGINAL PLANT
UPGRADED PLANT
WASTEMTER
SLUDGE
LOW-RATE
ANAEROBIC
DIGESTER
cw
DRYING
BEDS
ULTIMATE DISPOSAL
FINAL
CLAFIFiERS
SLUDGE THICKENERS
/CONVERTED
PRIMARY
\CLARIFIERS
GRIT REMOVAL
COMMINUTION
ANAEROBICf )
DIGESTER V7
DRYING
BEOS
ULTIMATE
DISPOSAL
CHLORINE CONTACT
TANK
HYDRASIEVE
NEW
PLASTIC MEDIA
ROUGHING FILTER
FINAL
CLARIFIERS
(ONE NEW)
CHLORINE CONTACT
TANK
-------
TABLE] 3-1
CASE HISTORY NO. 1 - PLANT OPERATING AND PERFORMANCE DATA
Description Before Upgrading After Upgrading
(1962)
Average Flow Rate, mgd 1.15 2.3
Primary Clarifier
Overflow Rate 1, gpd/sq ft 1,170
Hydrasieve Slot Size, in. 0.06
BOD Removal, percent 353 254
SS Removal, percent - 5 224
Plastic Media Roughing Filter
Hydraulic Loading6, gpm/sq ft 2.5
Organic Loading6, Ib BOD/day/1,000 cu ft - 520
Recirculation Ratio «2.0
BOD Removal, percent - 30
Trickling Filter (Stone Media)
Hydraulic Loading, mgd/acre 6.0 12.0
Organic Loading, Ib BOD/day/1,000 cu ft 56.2 87.0
Final Clarifiers
Overflow Rate, gpd/sq ft 1,170 750
Overall Plant Performance
BOD Removal, percent 83 85
SS Removal, percent ^ 84
Effluent BOD, mg/1 41 37
Effluent SS, mg/1 - 5 40
^ Based on average flow rate.
2 Primary clarifiers converted to gravity thickeners.
* Based on primary clarifier performance.
4 Based on hydrasieve performance only.
5 Performance data not available.
" Including recirculation.
13-3
-------
FIGURE 13-2
HYDRASIEVE SCREENING UNIT
,-1 y
COURTESY OF THE BAUER BROS. CO. - SPRINGFIELD, OHIO
13-4
-------
Water Pollution Board set a new final effluent BOD requirement of 20 mg/1, which Sedalia
could not meet with the existing facilities. The 1963 plant flow diagram is illustrated on
Figure 13-3.
The plant was upgraded to treat an average design flow of 2.5 mgd. The existing stone-media
filter was renovated to operate with plastic media. In addition, a second plastic-media filter
was constructed. The two plastic-media filters are operated in parallel with a total
recirculation ratio of 1.55. One additional primary clarifier and one additional secondary
clarifier were installed.
To remove additional BOD and SS, a shallow aerobic polishing lagoon was constructed after
the secondary clarifiers, and a vacuum filter was added to dewater the additional volume of
digested sludge. A flow diagram of the upgraded plant is also shown on Figure 13-3. Table
13-2 contains a summary of operating and performance data for the 1963 overloaded
period, along with the upgraded design criteria and actual operating and performance data
for the post-upgrading period. It should be noted that the effluent BOD was improved from
115 mg/1 to 11 mg/1 after upgrading, which is considerably below the 20 mg/1 requirement.
The capital costs of upgrading the plant were estimated at $3,055,000.
13.3 Case History No. 3 - Upgrading Using Polyelectrolyte Addition Before the Primary
Clarifier (5)
The Easterly Wastewater Treatment Works in Cleveland, Ohio, is a conventional activated
sludge plant whose dry-weather design flow is 123 mgd. In 1968, the plant was hydraulically
overloaded. Operating data from the 1968 control period are presented in Table 13-3. It is
interesting to note that the overflow rate of the primary clarifiers was 2,030 gpd/sq ft,
which is substantially above normally accepted values. As a result of the hydraulic
overloading, the overall BOD and SS removals were only 79 and 85 percent, respectively.
To improve the overall plant performance, a polyelectrolyte addition program was initiated.
An anionic polyelectrolyte, Purifloc A-23, was added at an average dosage of 0.21 mg/1.
Since proper polyelectrolyte dispersal and uniform mixing into the entire waste flow
constitute an extremely important aspect of the flocculation process, it was decided to add
the polyelectrolyte at the plant's two Venturi meters. These meters are located immediately
after the grit chamber and in front of the preaeration basin. Dye studies indicated that there
was a 7.5-minute travel time between the Venturi meters and the primary clarifiers. Six of
the 7.5 minutes were spent in the preaeration basin. The gentle agitation in the preaeration
basin provided adequate flocculation of wastewater solids.
13-5
-------
FIGURE 13-3
CASE HISTORY NO.2
COMPARISON OF ORIGINAL AND UPGRADED FLOW DIAGRAMS
LEGEND
- HASTEWATER
-- SLUDGE
OVERLOADED PLANT
OJ
ANAEROBIC
DIGESTION
ORYI
BED
NG
GRIT REMOVAL
COMMINUTION
r
4
A,
PRIMARY
"\JCLARIFIER
0
T
ULTIMATE
DISPOSAL
NOTE: CONVERTED TO USE PLASTIC MEDIA.
UPGRADED PLANT
r-
ANAEROBIC
DIGESTION
.__-_**
DRYING
BEDS
SECONDARY
CLARIFIER
I *-l
I I
NEW.A-
VACUUM
FILTER
ULTIMATE
DISPOSAL
EXISTING
TRICKLING
FILTER
REMOVAL
NUTION
PRIMARY
CLARIFIERS
(ONE NEW)
NEW
PLASTIC-
MEDIA
FILTER
SECONDARY
CLARIFIERS
(ONE NEW)
NEW SHALLOW
AEROBIC
POLISHING
LAGOON
-------
TABLE 13-2
CASE HISTORY NO. 2 - PLANT OPERATING, PERFORMANCE AND DESIGN DATA
1963 1969
Operating and Operating and
Performance Upgraded Performance
Description Data Design Data
Average Daily Flow, mgd 1.25 2.5 1.80
Raw Waste water BOD, mg/1 768 576 450
Primary Clarifiers
Overflow Rate1, gpd/sq ft 1,000 1,000 720
BOD Removal, percent 40 40 60
Trickling Filters
Hydraulic Loading^, mgd/acre 25 32 23
Organic Loading2, Ib BOD/day/1,000 cu ft 226 73 30
Recirculation Ratio 1.0 1.55 1.55
Final Clarifiers
Overflow Rate1 gpd/sq ft 755 755 545
Secondary BOD Removal, percent 75 93.2 86.8
Polishing Lagoon (Shallow Aerobic)
Maximum BOD Loading, Ib BOD/acre/day - 68 30
BOD Removal, percent - 12 54
Vacuum Filtration Rate, Ib/sq ft/hr - 5.0 - 3
Overall Plant Performance
BOD Removal, percent 85 96.5 97.7
Effluent BOD, mg/1 115 20 11
1 Based on average daily flow.
2 Including recirculation.
^ Lack of sufficient operating data.
13-7
-------
TABLE 13-3
CASE HISTORY NO. 3 - PLANT OPERATING AND PERFORMANCE DATA
1968 Polyelectrolyte
Control Demonstration
Description Period Period
Influent BOD, mg/1 104 67
Influent SS, mg/1 169 157
Primary Clarifier
Overflow Rate, gpd/sq ft 2,030 2,170
BOD Removal, percent 31 46
SS Removal, percent 31 51
Sludge Solids Concentration, percent 4.1 4.3
Sludge Volume Pumped, million gallons/month 5.0 6.8
Aeration Tank
MLSS, mg/1 1,670 1,602
Organic Loading, lb BOD/day/lb MLSS 0.48 0.29
Dissolved Oxygen, mg/1 3.2 3.8
Waste Activated Sludge Concentration, percent 2.4 2.0
Waste Activated Sludge Pumped,
million gallons/month 12.3 9.8
Overall Plant Performance
BOD Removal,, percent 79.1 83.4
SS Removal, percent 85.3 89.2
Effluent BOD, mg/1 21.8 11.1
Effluent SS, mg/1 24.8 17.0
A summary of the effectiveness of the poly electrolyte addition on plant performance as
compared to performance during the previous control period is also presented in Table 13-3.
The improvement in primary SS removal increased the volume of primary sludge from 5.0
to 6.8 million gallons per month. Overall plant performance was improved, as noted by the
reduced effluent BOD and SS concentrations. In addition to the increased treatment
efficiency, the polyelectrolyte addition was responsible for the following benefits to the
downstream activated sludge process:
1. A 20 percent volume decrease in waste activated sludge production
2. A 22 percent reduction in air supply requirements, resulting in a power cost
savings of over $3,300 per month
13-8
-------
3. An increase in aeration tank DO concentration from an average of 3.2 mg/1 to
3.8 mg/1.
An economic comparison was made between polyelectrolyte addition and providing
additional tankage to equal the performance of the flocculation system. The amortized cost
for the additional tankage was about $369,000 per year, while the chemical cost was
$158,000 per year, thereby indicating a considerable yearly savings in favor of the
polyelectrolyte alternative.
13.4 Case History No. 4 - Upgrading a Trickling Filter Plant by Adding Activated Sludge
Treatment and Pre- and Post-Chlorination
In 1967, the Livermore, California Wastewater Treatment Plant was upgraded to increase
plant capacity from 2.5 to 5.0 mgd and to provide a higher degree of treatment (6). All of
the existing treatment facilities were used in the upgraded treatment process. Flow diagrams
that show the type, number and arrangement of process units before and after upgrading are
presented on Figure 13-4.
The original plant was constructed in 1958 to provide secondary treatment for a domestic
wastewater flow of 2.5 mgd. Effluent standards in effect at that time required that the plant
produce an effluent that contained no more than 40 mg/1 of BOD and 40 mg/1 of SS. To
achieve this degree of treatment, preliminary and primary treatment facilities, a stone-media
trickling filter, a final clarifier and a 30-day polishing pond were provided. Waste sludge was
stabilized by anaerobic digestion and dewatered on sludge drying beds. This treatment
process produced an effluent that contained 45 to 50 mg/1 of BOD and 45 to 50 mg/1 of SS.
The upgraded plant was designed to produce an effluent that contained not more than
20 mg/1 of BOD, 20 mg/1 of SS and 1 mg/1 of grease. Bacteriological requirements limited
the average total coliform count to not more than 5 MPN per 100 ml over a five-day period.
To meet these design criteria, the plant facilities were expanded to provide activated sludge
treatment following first-stage trickling filtration along with pre- and post-chlorination. As
indicated on Figure 13-4, the existing grit chambers were adequate for the increased flow.
Sufficient primary clarification capacity at the higher flow was provided by converting the
existing final clarifier to a primary clarifier. An additional stone-media trickling filter was
constructed to provide for the increased hydraulic loading. To improve the degree of
treatment, two new aeration tanks, a new larger final clarifier and a new chlorination
contact tank were added to the treatment sequence. The polishing pond was converted to an
emergency holding basin from which any substandard effluent could be recycled through
the treatment plant. The sludge handling and treatment system was designed to return
excess waste activated sludge to the head end of the plant, and to anaerobically digest
combined primary and waste activated sludge. The existing sludge drying beds and two new
sludge lagoons were used for dewatering the digested sludge. A cost saving feature of the
upgraded plant is that trickling filter effluent is fed directly to the aeration tanks without
intermediate clarification.
13-9
-------
FIGURE 13-4
CASE HISTORY NO.4
COMPARISON OF ORIGINAL AND UPGRADED FLOW DIAGRAMS
EFFLUENT TO
POLISHING POND
DRV CAKE
TO LANDFILL
RAW
»ASTE*AT£R
BEFORE UPGRADING
RECIRCULATION
(1 EXISTING, I FINAL
CONVERTED TO PRIHARl),
i
EXISTING
AERATED
GRI T CHAMBERS
PRIMARY
CLARIFIERS
WASTE
SlUOGE
COMBINED
PRIHARY
AND
HASTE
SLUOGE
EFFLUENT
/EXISTING \
J ANAEROBIC )»
V DIGESTERS J
FILTRATE OR SUPERNATANT
AFTER UPGRADING
13-10
-------
Design conditions before and after upgrading are presented in Table 13-4. A summary of
treatment performance before and after upgrading is presented in Table 13-5.
As indicated in Table 13-5, the plant is now producing an effluent that well exceeds the
effluent standards for which it was designed. The effluent concentrations of BOD and SS
represent, respectively, 96.6 and 94.3 percent removals of the raw wastewater
concentrations of these two constituents.
The plant is presently achieving 99 percent oxidation of ammonia nitrogen. This is
important, not only because of the oxygen demand associated with ammonia nitrogen, but
also because the presence of ammonia compounds would require excessive chlorine dosages
to achieve the required high levels of disinfection.
The plan for upgrading requires that the existing digesters operate as high-rate units to
accommodate the increased quantity of sludge. Under present operating conditions, each
digester is loaded at an average rate of 0.22 Ib of dry solids/cu ft/day. At this loading rate,
approximately 54 percent reduction in volatile solids is achieved and the digested sludge
contains 1.6 percent solids by weight. This low concentration of solids coupled with the
lack of mechanical cleaning facilities has caused some difficulty in attaining optimum use of
the sludge drying beds. Accordingly, most of the digested sludge is dewatered in the sludge
lagoons. The recycled flow from the sludge lagoons has caused some difficulties in plant
operation. For these reasons, various plans for upgrading the sludge treatment and disposal
system are now under consideration (7).
The cost of upgrading the plant was $1,950,000, and the annual operating costs for 1970
were estimated at $227,840, exclusive of administrative costs.
13.5 Case History No. 5 - Upgrading a Primary Treatment Plant to Provide Tertiary
Treatment (8)
The Central Contra Costa Sanitary District of Walnut Creek, California, undertook
plant-scale studies to determine the best method of upgrading a primary plant to produce an
effluent that would be acceptable for industrial cooling water use. The findings of this
investigation indicated that if the existing plant facilities were modified and expanded to
provide carbonaceous and nitrogen oxidation, denitrification and filtration, the effluent
would be acceptable for the intended use. Flow diagrams for the existing and proposed
plant are shown on Figure 13-5.
Operating conditions for the existing plant and proposed design parameters for the upgraded
plant are shown in Table 13-6. Most of the proposed design parameters are based on the
results obtained from the plant-scale studies.
13-11
-------
TABLE 13-4
CASE HISTORY NO. 4 - PLANT OPERATING AND DESIGN CONDITIONS
Description
Pretreatment Facilities Capacity, mgd
Preaeration Tanks
Number of Units
Detention Time, hr
Air Supplied per tank, cfm
Hydraulic Capacity, mgd
Primary Clarifiers
Number of Units
Overflow Rate, gpd/sq ft^
Hydraulic Capacity
Trickling Filters
Number of Units
Hydraulic Loading, mgd/acre
Organic Loading, Ib BOD/day/acre-ft
Recirculation Ratio
Aeration Tanks
Number of Units
Detention Time, hr
Volumetric Loading, Ib BOD/day/1,000 cu ft
Sludge Recycle, percent1
Final Clarifier
Number of Units
Detention Time, hr
Overflow Rate, gpd/sq ftl
Polishing Pond
Number of Units
Detention Time, days
Organic Loading, Ib BOD/day/acre-ft
Operating
Conditions
Before
Upgrading
2.5
2
0.6l
200
10
2
1,050
10
1
2.0
822
1
30
28
Design
Conditions
After
Upgrading
5.0
2
0.62
150
10
2
1,050
10
2
34.5
4,600
1.5 to 3.0
2
34.5
4,400
1.5 to 3.0
2
5.2
28
10 to 100
1
2.75
787
13-12
-------
TABLE 13-4 (Continued)
Operating
Conditions
Before
Upgrading
2
0.07
Description
Anaerobic Digesters
Number of Units
Loading, Ib dry solids/cu ft/day
Chlorine Contact Tank
Number of Units
Detention Time, hr
Sludge Drying Beds
Number of Units
Total Area, sq ft
Digested Sludge Lagoons
Number of Units
Volume each, 1,000 cu ft
1 Only one tank in service.
2 Both tanks in service.
3 Average dry weather flow.
TABLE 13-5
SUMMARY OF TREATMENT PERFORMANCE FOR CASE HISTORY NO. 4
4
22,400
Design
Conditions
After
Upgrading
2
0.22
1
1.0
4
22,400
2
320
Parameter
Average Flow, mgd
Effluent BOD, mg/1
Effluent SS, mg/1
NHg-N, mg/1
N03-N, mg/1
Conforms, MPN/100 ml
Grease, mg/1
Measured Performance
Before Upgrading
2.5
45-50
45-50
_ 1
_ 1
Measured Performance
After Upgrading^
3.37
7.3
13
0.14
21.5
2.5
0.23
1 Before upgrading, the plant produced no significant nitrification.
^ Monthly average.
13-13
-------
co
FIGURE 13-5
CASE HISTORY NO.5
COMPARISON OF ORIGINAL AND UPGRADED FLOW DIAGRAMS
SCREENED
RA* ».
WASTEWATER
AERATED
GRIT
CHAMBER
PRIMARY
CURIFI ERS
i
CHLORINE
CONTACT
TANK
PRIMARY
SLUDGE
/ ANAEROBIC\
( DIGESTORS f"
* TO SLUDGE LAGO
EFFLUENT
BEFORE UPGRADING
EFFLUENT TO
^-INDUSTRIAL
REUSE
> EFFLUENT TO
RECEIVING WATER
WASTE SLUDGE
AFTER UPGRADING
-------
TABLE 13-6
CASE HISTORY NO. 5 - PLANT OPERATING AND DESIGN CONDITIONS
Description
Preliminary Treatment (Bar Screen, Raw
Wastewater Pumps) Average Capacity, mgd
Preaeration-Grit Removal Tanks
Detention Time (ADWF1), hr
Design Flow, mgd
Primary Clarifiers
Number of Units
Detention Time (ADWF), hr
Overflow Rate (ADWF), gpd/sq ft
Aeration and Nitrification Tanks
Number of Units
Detention Time, hr
Volumetric Loading, Ib BOD/day/1,000 cu ft
Intermediate Clarifiers
Number of Units
Detention Time (ADWF), hr
Overflow Rate (ADWF), gpd/sq ft
Overflow Rate (peak flow), gpd/sq ft
Denitrification Tanks
Number of Units
Detention Time (ADWF), minutes
Final Clarifiers
Number of Units
Detention Time (ADWF), hr
Overflow Rate (ADWF), gpd/sq ft
Overflow Rate (peak flow), gpd/sq ft
Effluent Polishing Filters
Number of Units
Filtration Rate (ADWF), gpm/sq ft
Backwash Water Rate, gpm/sq ft
Maximum Rate
Minimum Rate
13-15
Operating
Conditions
Before
Upgrading
31
0.5
31
1.5-1.62
1,040-1,200
Design
Conditions
After
Upgrading
30
0.37
30
4
2.4
720
2
6.8
27.3
4
5
788
2,000
2
103
4
5
788
2,000
4
4.0
25
10
-------
TABLE 13-6 (Continued)
Description
Sludge Digesters
Number of Units
SRT, days
Loading, Ib dry solids/cu ft/day
Sludge Dewatering and Incineration
Centrifuges, Number of Units
Rate, gpm
Polymer dosage, Ib/ton of dry solids
Incinerators, Number of Units
Furnace Capacity, Ib/hr
Sludge Drying Beds
Number of Units
Total Area, sq ft
Sludge Lagoons
Number of Units
Capacity, 1,000 Ib dry solids
Chlorination, Ib/day
Operating
Conditions
Before
Upgrading
4
25
0.15
4
20,000
2
15,600
18,000
Design
Conditions
After
Upgrading
2
100
4-5
3
51,200
ADWF = Average dry weather flow.
Varies from tank to tank.
The original primary treatment plant was constructed in 1948, with expansions in 1957 and
1964 to increase the plant capacity to 31 mgd. Pretreatment including pre-chlorination,
screening, and grit removal, followed by primary sedimentation and chlorine disinfection
produced an effluent with 144 mg/1 of BOD. Sludge produced in the primary tanks was
anaerobically digested and then discharged to sludge lagoons and drying beds.
The upgraded plant is designed to produce an effluent that is suitable for either industrial
reuse or discharge into receiving waters. As indicated on Figure 13-5, the design calls for the
13-16
-------
addition of lime to the wastewater flow, after screening. The plant-scale field studies
indicated that when lime is mixed with the wastewater in the aerated grit chambers, it
combines with phosphorus and other substances to form a floe that improves the
coagulation of solids. As a result, increased BOD, SS and phosphorus removals may be
expected in the primary clarifiers. Effluent from the primary tanks will be pumped to the
aeration-nitrification tanks, where carbonaceous organics and ammonia nitrogen will be
simultaneously oxidized. The increased BOD removal accomplished in the primary clarifiers
due to lime addition lowers the organic loading on the aeration tanks and is the key
upgrading feature that will permit nitrification in a single-stage activated sludge system. The
previously mentioned plant-scale studies also indicated that recarbonation of the
lime-treated primary effluent prior to entering the aeration tanks was not necessary when
the lime clarification process was operated at a pH of 11.0 or less, due to the C02 and nitric
acid produced in the nitrification reactor.
Nitrified effluent from the intermediate clarifiers will be fed to a stirred anaerobic reactor
for biological denitrification. Methanol will be added as an energy source for the
denitrifying organisms. Following denitrification, the wastewater passes through a short
detention aerobic polishing reactor for oxidizing residual methanol and for stripping
nitrogen gas formed prior to final clarification. After final clarification, the effluent will be
chlorinated and the portion required for industrial use pumped to dual-media polishing
filters and final chlorine disinfection. Any remaining effluent will be discharged to the
receiving water.
The sludge handling and treatment system will be upgraded and expanded to accommodate
the increased quantities of sludge. Waste biological sludge will be returned to the plant
influent ahead of the point of lime addition and settled in the primary tanks together with
the chemical-primary sludge. The existing sludge digesters will be modified to serve as
holding and mixing tanks for the combined sludge. The blended sludge will be classified and
dewatered by two-stage centrifugation prior to incineration in multiple-hearth furnaces.
The upgraded plant will provide for an average dry weather flow capacity of 45 mgd
through the primary treatment units, and 30 mgd through the remainder of the plant. The
design and layout for this upgrading were planned such that subsequent plant expansions
will ultimately provide the capacity to treat a future average dry weather flow of 120 mgd.
All hydraulic structures for the primary and secondary treatment facilities are designed to
carry the ultimate maximum wet weather flow of 300 mgd.
Design criteria for the existing and the upgraded plant are presented in Table 13-6. A
summary of the treatment plant performance before upgrading and the anticipated
performance after upgrading is given in Table 13-7. Costs for the upgrading are presented in
Table 13-8.
13-17
-------
TABLE 13-7
SUMMARY OF TREATMENT PERFORMANCE FOR CASE HISTORY NO. 5
Measured Performance Anticipated Performance
Parameter Before Upgrading After Upgrading 1
Average Design Flow, mgd 31 30
Effluent BOD, mg/1 103-137 2
Effluent SS, mg/1 57-64 1
Effluent Total Nitrogen, mg/1 - 2
Effluent Total Phosphorus, mg/1 0.2
After effluent polishing for industrial reuse.
TABLE 13-8
ESTIMATED COSTS FOR CONSTRUCTION FOR CASE HISTORY NO. 5
Description Estimated Cost
Preliminary Treatment Works $ 680,000
Preaeration-Grit Removal Structure 540,000
Primary Clarifiers 920,000
Aeration and Nitrification Tanks 4,480,000
Aeration Blower Building and Equipment 1,620,000
Intermediate Clarifiers 1,710,000
Denitrification Tanks 860,000
Final Clarifiers 1,670,000
Primary Effluent and Final Effluent Pumping Structure 3,430,000
Incineration and Dewatering Building and Equipment 4,000,000
Chemical Storage Area 650,000
Dual-Media Filters 1,900,000
Effluent Storage, Chlorination and Pumping Facilities 2,380,000
Administration Building Laboratory 1,330,000
Maintenance Building 610,000
Utility Tunnel 950,000
Outside Piping and Site Development 1,400,000
Total Construction Cost $29,130,000
13-18
-------
13.6 Case History No. 6 - Upgrading a Trickling Filter Plant in Stages to an Activated
Sludge Plant with Roughing Filters (6)
The original South Buffalo Creek Wastewater Treatment Plant in Greensboro, North
Carolina, was built in 1931 for an average flow of 3.25 mgd. This facility provided primary
clarification and secondary treatment using fixed-nozzle square-tank trickling filters. Sludge
was digested anaerobically and dewatered on sludge drying beds.
The existing plant was first upgraded in 1957 when treatment efficiency became inadequate
due to an increase in average flow to 4.0 mgd, including 1.5 mgd of industrial wastes. The
upgrading included replacement of the screening and grit removal units, expansion of
primary clarification, final clarification and digestion facilities, installation of new circular
trickling filters and addition of chlorine disinfection facilities. Replacement of the drying
beds with a vacuum filter installation was also required due to the additional sludge
produced by the upgraded treatment system. The treatment provided by this upgraded
system resulted in BOD and SS removals of 89 percent and 84 percent, respectively.
Increases in the industrial BOD influent loads beyond those anticipated in the 1957 design,
required further plant upgradings in 1964 and 1970. In each of these upgradings, one of the
two original fixed-nozzle trickling filters was converted into two aeration basins for
activated sludge treatment. Waste activated sludge was returned to the plant influent and
settled with the primary sludge. In 1970, digestion of the combined sludge was discontinued
and trucking of the dewatered raw sludge to an incinerator replaced landfilling for ultimate
disposal.
Current upgrading (presently under construction) to increase plant capacity will include
chemical addition to the raw wastewater for phosphorus removal, separate flotation
thickening of waste activated sludge to reduce the loading on the dewatering facilities and
effluent polishing including postaeration and dual-media filtration. Under this current
upgrading, the plant will provide 98 percent BOD and SS removals for an ultimate dry
weather flow of 10.7 mgd.
A flow diagram and design data for the original plant and each subsequent upgrading are
shown on Figure 13-6 and Table 13-9, respectively. Plant performance data before and after
each upgrading are shown in Table 13-10.
Capital costs for each stage of the upgrading are presented in Table 13-11.
13-19
-------
FIGURE 13-6
COMPARISON OF ORIGINAL AND UPGRADED FLOW DIAGRAMS
INFLUENT
WASTEWATER
SLUDGE
DRYING
BEDS
1
TO LANDFILL
t
/AN
\ "
PRIM
CLARI
\
r
AEROBIC \
IGESTER J*
'
FIXED-
ARY ^ NOZZLE .. FINAL
FIER * TRICKLING ' CLARIFIER
FILTERS
^ WASTE SLUDGE ''
EFFLUENT
COMBINED PRIMARY
AND WASTE SLUDGE
OR GINAL PLANT f 931)
EXISTING
FIXEO-
* NOZZLE
TRICKLING
FILTERS
(1 EXISTING,
1 NEW)-^ ^> ^
' / NEW \
INFLUENT PRIMARY / TR1CnL1HG \
""ilc""":" " ' CLARIFIERS T FILTERS /
^^ WASTE SLUDGE
PRIMARY OR
/ \ SLUDGE
NEW VACUUM / ANAEROBIC \^ ir
FILTER I OIGESTERSr WASTE SLUDGE
4 \|l NEW (PRIMARY)
SLUDGE CAKE }l EXISTING (SECONDARY)
TO
LANDFILL
NEW
CLARIFIERS
(CONVERTED
CHLORINE EXISTING
A - Jtl\u M n i FINAL
T,u «!«.»
4
EFFLUENT
PLANT AFTER 1957 UPGRADING
13-20
-------
FIGURE 13-6
(CONTINUED]
INFLUENT
KASTEWATER
(CONVERTED
FIXED-NOZZLE
TRICKLING FILTERS)
SLUDGE
AERATION
TANKS
VERTED
EXISTING
FINAL
CLARIFIERS
^-y"
1
EXI STING
CHLORINE
CONTACT
TANK
ANAEROBIC
DlGESTERS)
SLUDGE CAKE
TU INCINERATOR
»ASTE SLUDGE
PLANT AFTER 1970 UPGRADING
OPTIONAL
EQUALIZATION
INFLOENT _
KASTEHATER
NEW
PREAERATION
TANK
EXISTING
PRIMARY
CLARIFI ERS
v^ ./
EFFLUENT
SLUDGE CAKE
TO INCINERATOR
PLANT AFTER CURRENT UPGRADING
13-21
-------
TABLE 13-9
CASE HISTORY NO. 6 - PLANT DESIGN CONDITIONS
co
Description
Design Average Flow, mgd
Preaeration Tanks
Detention Time (DAFl), minutes
Primary Settling Tanks
Detention Time (DAF), hr
Overflow Rate (DAF), gpd/sq ft
Secondary Treatment
Square Trickling Filters
Organic Loading (DAF),
Ib BOD/day/1,000 cu ft
Hydraulic Loading (DAF), mgd/acre^
Round Trickling Filters
Organic Loading (DAF),
lbBOD/day/l,OOOcuft
Hydraulic Loading (DAF), mgd/acre^
Aeration Tanks
Detention Time (DAF), hr
Final Clarifiers
Detention Time (DAF), hr
Overflow Rate (DAF), gpd/sq ft
Chlorination Tanks
Contact Time (DAF) minutes
Original Plant
1931
3.25
2.7
670
12
1.8
1.4
1,300
1957
8.0
2.1
810
15
2.2
31
4.6
3.5
460
1964
8.0
2.1
810
62
9.2
7.3
3.5
460
1970
8.0
2.1
810
62
9.2
14.7
3.5
460
Current
Upgrading
10.7
14.8
1.6
1,080
83
12.4
11.0
2.6
620
34
34
34
-------
TABLE 13-9 (Continued)
to
CO
Description
Sludge Thickening
Flotation Unit Surface Area, sq ft
Sludge Digestion (Number of Units)
60 ft diameter x 23 ft SWD anaerobic
80 ft diameter x 30 ft SWD anaerobic
Sludge Dewatering
Sludge Drying Beds, sq ft
Vacuum Filter Surface Area, sq ft
Ultimate Disposal
Original Plant
1931
32,600
Landfill
1957
430
Landfill
1964
1970
Current
Upgrading
250
430
Landfill
430
Incineration
430
Incineration
DAF = Design Average Flow.
Does not include recirculation.
-------
TABLE 13-10
SUMMARY OF TREATMENT PERFORMANCE FOR CASE HISTORY NO. 6
1957
1964
1970
co
fe
Average Flow, mgd
Peak Flow, mgd
BOD
Raw Waste water, mg/1
Removal, percent
SS
Raw Wastewater, mg/1
Removal, percent
Before
Upgrading
3.7
10.0
380
60- *
350
60
After
Upgrading
4.5
11.5
310
90
240
85
Before
Upgrading
4.9
11.5
290
70
250
75
After
Upgrading
6.6
16.0
300
90
260
80
Before
Upgrading
6.8
16.0
390
85
290
75
After
Upgrading
8.9
18.0
390
90
300
75
Current
Upgrading
10.7
21.0
400
>98
300
>98
-------
TABLE 13-11
CAPITAL COSTS OF UPGRADING FOR CASE HISTORY NO. 6
Capital Cost
Date of
Upgrading
1957
1964
1970
Current
Design
Item
Screening and Degritting
Rehabilitation of Pump and
Control Building
Primary Clarification
Secondary Process
Final Clarification
Chlorination Facilities
Sludge Handling
Miscellaneous
Total
Modifications to Secondary Process
Modifications to Secondary Process
Renovation of Screens and
Degritting Equipment
Preaeration and Odor Control
Enclosure of Primary Tanks and
Odor Control
Phosphorus Removal Facilities
Effluent Filters and Aerators
Sludge Handling
Miscellaneous
Total
Cost at Time
of Upgrading
$ 45,000
150,000
65,000
330,000
135,000
25,000
300,000
150,000
$1,200,000
200,000
100,000
1972
Cost
$ 110,000
365,000
160,000
800,000
330,000
60,000
730,000
365,000
$2,920,000
367,000
133,000
60,000
160,000
200,000
75,000
450,000
250,000
55,000
$1,250,000
13.7 Case History No. 7 - Upgrading by Optimization of Aeration Tank-Clarifier
Relationship (9)
The Coldwater Creek Wastewater Treatment Plant in St. Louis, Missouri, is an activated
sludge plant designed to treat a flow of 25 mgd. The plant facilities include six aeration
tanks and four final clarifiers; however, at the time of upgrading, only three of the aeration
tanks were being used to treat a flow of 13.2 mgd. Under these operating conditions, the
plant efficiency was poor with BOD and SS removals of 73 and 46 percent, respectively.
13-25
-------
Table 13-12 summarizes operating data before and after optimization of the aeration
tank-clarifier relationship. The flow diagrams, presented on Figure 13-7, depict the changes
in operation that were made. Analysis of the data indicates that before the modifications,
the return sludge rate and the resulting MLSS concentration in the aeration tanks were both
high. These conditions imposed a solids loading rate on the final clarifiers in excess of the
recommended range (20-30 Ib/day/sq ft) shown in Table 6-2, even though the overflow rate
was within acceptable limits. As a result, poor clarifier performance lowered the overall
BOD and SS removals normally expected in an activated sludge system.
TABLE 13-12
CASE HISTORY NO. 7 - OPERATING DATA
Before Operating After Operating
Description Modifications Modifications
Average Flow, mgd 13.2 14.5
Aeration Tanks
Tanks in Service 3 4
MLSS, mg/1 7,200 3,400
Return Sludge Rate, percent
of Average Flow 93 30
Volumetric Loading, IbBOD/day/l,000 cu ft 44.7 31.5
Organic Loading, Ib BOD/day/ Ib MLSS 0.10 0.15
Final Clarifiers
Tanks in Service 4 3
Overflow Rate at Average Flow, gpd/sq ft 515 755
Solids Loading Rate at Average Flow 59.8 27.8
To improve plant performance, a fourth aeration tank was put in service. A lower return
sludge rate was used and the MLSS concentration in the aeration tanks was decreased by
over 50 percent. With these operational modifications, it was possible to maintain both the
final clarifier solids loading rate and the overflow rate within their recommended limits with
only three final clarifiers in service. The improvement in treatment plant efficiency resulting
from the proper balancing of the aeration tanks and final clarifiers is shown in Table 13-13.
13-26
-------
FIGURE 13-7
CASE HISTORY NO.7
COMPARISON OF ORIGINAL AND MODIFIED FLOW DIAGRAMS
4 PRIMARY CLARIFIERS
(ALL IN SERVICE)
6 AERATION TANKS
(3 IN SERVICE)
4 FINAL CLARIFIERS
(ALL IN SERVICE)
RAW
WASTEWATER i
r~*
->
h-
1 P
^
^
LU
CJ9
a
__i
oc.
UJ
oc
r >
1 ^
->|
fc
fe
fc-
^
^>
^
^
^
EFFLUENT
EXCESS SLUDGE
FLOW DIAGRAM BEFORE OPERATING MODIFICATION
PRIMARY CLARIFIERS 6 AERATION TANKS
(ALL IN SERVICE) (4 IN SERVICE)
4 FINAL CLARIFIERS
(3 IN SERVICE)
RAW
HASTEWATER *
t w'
1 ^
- -w
^
'
^
i *
*
cr>
CO
*
LU
0=
-fc
»,
h
^
k.
k
^
EFFLUENT
EXCESS SLUDGE
FLOW DIAGRAM AFTER PLANT MODIFICATION
13-27
-------
TABLE 13-13
CASE HISTORY NO. 7 - PERFORMANCE DATA
Description
Average Wastewater Flow, mgd
BOD
Raw Wastewater, mg/1
Primary Effluent, mg/1
Final Effluent, mg/1
Overall Plant Removal, percent
SS
Raw Wastewater, mg/1
Primary Effluent, mg/1
Final Effluent, mg/1
Overall Plant Removal, percent
Before Modification
of Flow Pattern
13.2
150
152
40
73
173
155
92
47
After Modification
of Flow Pattern
14.5
162
130
9l
94
198
142
161
92
1 One week average.
13.8 Case History No. 8 - Upgrading by Optimization of Aeration Tank-Clarifier
Relationship (9)
The Sioux Falls, South Dakota Wastewater Treatment Plant is a 9.5-mgd activated sludge
plant. Approximately 3.5 mgd of the flow is industrial wastes (meat-packing wastes with a
BOD of 2,500 mg/1). The industrial wastes are pretreated by clarification and high-rate
trickling filtration. The pretreated industrial wastes are then added to the main wastewater
flow which is primarily domestic, and the combined flow is treated by primary clarification
followed by an activated sludge process. A flow diagram of the major treatment processes is
presented on Figure 13-8.
During most of the year, the plant provided excellent treatment. However, during the fall,
the treatment plant is subjected to high industrial loads that significantly lowered plant
efficiency prior to optimization of the aeration tank-clarifier relationship. Table 13-14
shows operating and performance data for the months of October and November, the period
when loads are highest. The aeration tank loadings during 1967 were extremely high, while
13-28
-------
the solids loading rate on the final clarifiers was low. The unbalanced loadings on the
aeration tanks and final clarifiers were attributed to the relatively low return sludge rate and
the resulting low MLSS concentration.
PRETREATMENT
CLARIFIERS
INDUSTRIAL
WASTEHfATET
DOMESTIC
WASTEWATER
FIGURE 13-8
CASE HISTORY NO.8
FLOW DIAGRAM
PRETREATMENT
HIGH-RATE
TRICKLING FILTER
EFFLUENT
PRIMARY
CLARIFIERS
AERATION
TANKS
FINAL
CLARIFIERS
To improve treatment plant efficiency during the peak load season, it was decided to lower
the organic loading on the aeration tanks by increasing the return sludge rate, thereby
raising the MLSS concentration. The success of this operational modification was dependent
on the ability of the final clarifiers to satisfactorily handle the increased solids loading.
Table 13-14 also presents operating and performance data for October and November, 1968,
after implementation of the modified operational mode. These data indicate that the final
clarifier solids loading rate was substantially raised, although still at an acceptable level,
without affecting the overflow rate. The improvements to treatment plant efficiency are
evident in Table 13-14.
13-29
-------
TABLE 13-14
CASE HISTORY NO. 8 - OPERATING AND PERFORMANCE DATA
1967 1968
Before Modification After Modification
Description of Operating Mode of Operating Mode
Average Wastewater Flow, mgd 10.0 9.8
Aeration Tanks
MLSS, mg/1 1,000 3,000
Return Sludge Rate,
percent of Average Flow 30 90
Volumetric Loading,
Ib BOD/day/1,000 cu ft 115 117
Organic Loading,
Ib BOD/day/lb MLSS 1.84 0.63
Final Clarifiers
Overflow Rate, gpd/sq ft 660 650
Solids Loading Rate,
Ib/day/sqft 7 31
Final Effluent1
BOD, mg/1 29.5 20.0
SS, mg/1 34.5 13.5
Fifty percent of the time effluent quality was equal to or less than values shown.
13.9 Case History No. 9 - Upgrading a Modified Aeration System for Nutrient
Removal (10)
In 1969, regulatory agencies established more stringent effluent standards for treatment
plants that discharge into the Potomac River in the vicinity of Washington, D. C. These
higher standards required upgrading the Washington, D. C. Blue Plains Plant to provide
phosphorus and nitrogen removal as well as improved BOD and SS removals.
In 1969, very little performance data were available on the alternative phosphorus and
nitrogen removal methods that might be used in this upgrading situation. Fortunately,
through the cooperation of the Joint EPA-DC Pilot Plant, it was possible to pilot and
evaluate several alternative nutrient removal treatment sequences. Based on these studies,
13-30
-------
two-point addition of a metal salt was selected for phosphorus removal, and it was
determined that nitrogen removal would be best achieved through biological nitrification
and denitrification processes. The pilot studies also indicated that to consistently meet the
established effluent standards, multimedia filtration would be required. Anticipated
performance data for the upgraded plant are presented in Table 13-15. Figures 13-9, 13-10,
and 13-11 show, respectively, the flow diagrams for the primary and secondary systems,
nitrification and denitrification systems and filtration and disinfection systems of the
upgraded plant.
The existing secondary system consists of four aeration tanks and 12 sedimentation units.
To handle the anticipated increase in plant design flow from 240 mgd to 309 mgd, the
existing secondary system will be upgraded with two additional aeration tanks and 12
additional final sedimentation tanks. The aeration tanks are designed for a volumetric
loading of 120 Ib BOD/day/1,000 cu ft, an organic loading of 2.4 Ib BOD/day/lb MLSS and
a MLSS concentration of 1,300 mg/1 (10). Since the increased design loadings require more
air per unit volume than the existing aeration system can deliver, the existing aeration
system will be expanded. This system will be modified from a coarse-bubble, spiral-roll
system to a coarse-bubble, spread-pattern to improve oxygen transfer efficiency. The
secondary system air capacity has been designed to provide 0.54 Ib C>2/lb BOD removed
(10).
Alum or ferric chloride will be added to the mixed liquor of the secondary system and is
expected to remove approximately 70 percent of the phosphorus contained in the plant
influent. To remove most of the remaining phosphorus, metallic salts will also be added to
the nitrogen release tanks. The addition of metallic salts to the secondary system is also
expected to improve the BOD removal in the secondary system from 75 percent to
85 percent. This will ensure a secondary effluent BOD concentration of less than 40 mg/1,
which was found during pilot testing to be desirable for successful nitrification in the second
stage.
Biological nitrification facilities are designed for oxidation of 0.066 Ib NH3~N/day/lb
MLVSS at minimum wastewater temperatures and a MLVSS concentration of 1,700 mg/1.
At the stoichiometric oxygen requirement of 4.6 Ib O2/lb NHgN oxidized, 120-75 hp
turbine aerators are required (10). Maximum air supply to the turbines will be 88,000 cfm.
The turbines were selected in this instance because, due to the limitations of the site, the
nitrification tanks are designed to have depths of 30 feet to obtain the required volume. The
turbines will provide adequate mixing to this depth and are capable of supplying a wide
range of oxygen to the system as required by varying NHgN influent loads and varying
wastewater temperatures. Lime will be added to the nitrification reactor to maintain the
necessary pH for nitrification. The nitrification sedimentation tanks are designed for average
and peak hydraulic and solids loadings of 580 and 1,210 gpd/sq ft and 17.4 and
36.6 Ib/sq ft/day, respectively. The sludge return system is designed to provide return of
13-31
-------
TABLE 13-15
CASE HISTORY NO. 9 - ANTICIPATED PERFORMANCE AFTER UPGRADING
co
BOD, mg/1
Total Phosphorus, mg/1
Nitrogen:
Organic-N, mg/l
NH3-N, mg/1
NO2 + NOa-N, mg/1
Total N, mg/1
Grit Chamber
Effluent
206
8.4
8.6
13.7
0
22.3
Secondary
Effluent1
35
2.0
3.0
14.8
0.2
18.0
Nitrification
Effluent
10
1.0
1.0
1.5
11.1
13.6
Denitri-
fication
Effluent
6
0.5
1.0
1.0
1.0
3.0
Filtration
Effluent
4
0.2
0.5
1.0
0.5
2.0
Effluent
Standard
5
0.22
2.4
With ferric chloride or alum addition.
-------
FIGURE 13-9
UPGRADING A MODIFIED AERATION SYSTEM FOR NUTRIENT REMOVAL
FLOW DIAGRAM
PRIMARY AND SECONDARY SYSTEMS (10)
OJ
oo
w
pa
ANACOSTIA 1 '
FORCE MAIN
mil
RAW T^|~PUMP "I Jnf*
WASTEWATER | >ift''^Np ~+
(5
Z _
N 3
^ 0
< Z
o 8
u*
2 !
"i ,_i
WASTEWATER^' 5TATIONJ * <"<"
LEGEND
CONTINUOUS INTERMITT
FLOW FLOW
(INC. AIR)
PSL PRIMARY SLUDGE
RSL RETURN SLUDGE
WSL WASTE SLUDGE
ALT.
SPENT
| WASHWATER
1
| FERRIC OR ALUM 1 !
: j|
m
\ DISINFECTION | EXCESS FLOW TO RIVER
ATRl ! 1
! .-- ~ -v JSE<
1 / \ TL"
ERATED ~|ir%*/ PRIMARY \Tl' ' f
CHAMBERlT T\j SEDIMENTATION |U- ' '
1 r\ TANKS 1
v^^x
r" SLUDGE
GRAVITY RECYCLE >SE<
^ THICKENERS
1 GRAVITY ' ^
pH THICKENING ,
1 j OVERFLOW j
^^^^^^^^^^^
EXIST k Jj EXIST \ 1
CBATcn 1 V PRIMARY 1 ISE(
ECRH"MDBER[T^;SEOI^TKASTIONI N
ENT r^- ["PCJLYMTR"
fTEl RiN THK
NITRIFICATION
j INFLUENT^
1
I
lATTlcf
FILTERS
1 OR
OUTFALL
L ^
ALT TO
ITRIFICATION l_ .
OR T"^
NITRIFICATION j
KENERS
-------
FIGURE 13-10
UPGRADING A MODIFIED AERATION SYSTEM FOR NUTRIENT REMOVAL
FLOW DIAGRAM
NITRIFICATION AND DENITRIFICATION SYSTEMS (10)
co
co
LIME
1
SECONDARY
EFFLUENT I
i
i
^
POLYMER AIR METHANOL
i i t
NITRIFICATION! 1
^ REACTORS l_^
NITRIFICATION! | pl.up ! I
SED. BASINS LJ^^U,
III 1:
t !'
«* i 1
t ' ^
ALT. WSL FROM
SECONDARY
OR
DENITRIFICATION
FERR
SPENT ...
WASHWATER AIR
1 '
DENITRIFICATION li r NIT*
REACTORS Uk yAL
COR ALUM
1 ' POLYMER
f
3AGSEEN JDEN''R'F'CA''?N| TO MULTI-MEDIA
MKi *| --CD. BA.IN, 1 F|ITERS »
t
i RSL
rWSl j JAITJ ^ J
^ "-1
I t
1
ALT WSL FROM !
SECONDARY
OR
NITRIFICATION
WSL W
TO
FLOTATION THICKENERS
AwSL
LEGEND
CONTINUOUS INTERMITTENT
FLOW FLOW
"iSTFWATFR _^i,__>. . i
(INC. AIR)
RSL RETURN SLUDGE
WSL WASTE SLUDGE
-------
FIGURE 13-11
UPGRADING A MODIFIED AERATION SYSTEM FOR NUTRIFENT REMOVAL
FLOW DIAGRAM
FILTRATION AND DISINFECTION SYSTEMS (10)
SPENT WASHWATER
TO NITROGEN RELEASE
co
co
cn
LEGEND
CONTINUOUS INTERMITTENT
FLOW FLOW
FLUSHING. SERVICE AND
DILUTION WATER
WASTEWATER
CHEMICALS
-------
40 percent of peak flow. However, the system will normally be operated to return
30 percent of the average flow. Continuous monitoring of the DO content of the
nitrification effluent will be provided to ensure that the influent DO to the denitrification
system is minimized.
The biological denitrification system will include reactors, nitrogen release tanks and
sedimentation tanks. The reactors have been designed for removal of 0.0425 Ib
N03-N/day/lb MLVSS at a design MLVSS concentration of 2,100 mg/1, with up to 4.5 Ib
methanol added/lb NOgN applied (10). The reactors will be equipped with 48-75 hp
mixers, and will be covered but not airtight.
The nitrogen release tanks will serve three functions: (1) to strip supersaturated nitrogen
gas, (2) to provide mixing for second-stage metal salt addition for residual phosphorus
removal and (3) to provide an aeration zone for removal of excess methanol. These tanks
will furnish a 23-minute detention period at average flow.
The denitrification sedimentation tanks are designed for hydraulic loadings of 670 gpd/sq ft
at average flow, and 1,410 gpd/sq ft at peak flow. Solids loadings are 25.6 and
54.0 Ib/sq ft/day at average and peak flows, respectively.
The 36 multimedia filters are designed for filtration rates of 3.0 gpm/sq ft at average flow
and 6.2 gpm/sq ft at peak flow. Backwashing will occur at normal intervals of 24 hours at a
rate of 25 gpm/sq ft. The backwash water will be equalized in conduits and may be returned
upstream of either the secondary reactors or the nitrogen release tanks.
Provision has been made to chlorinate either upstream or downstream of the filters with 24
minutes detention provided in contact tanks following the filters.
Sludge processing facilities will include gravity thickening of primary sludge, flotation
thickening of secondary and advanced treatment sludges, vacuum filtration and sludge
incineration.
13.10 References
1. Wittenmyer, J.D., and Sak, J.G., Plastic Media Roughing Filter Provides Most
Economical Plant Expansion. Presented at the Ohio Water Pollution Control
Association Conference (June 15, 1967).
2. Wittenmyer, J.D., A Look at the Future Now. Presented at the Ohio Water Pollution
Control Association Conference (June 20, 1969).
3. Private communication with J.D. Wittenmyer, Vice President, Ohio Suburban Water
Company, Dayton, Ohio (January 22, 1971).
13-36
-------
4. Burns & McDonnel Engineering Company. Report on Sewage Treatment Plant and
Sanitary Sewer Improvements for Sedalia, Missouri (1963).
5. Wirts, J.J., The Use of Organic Polyelectrolyte for Operational Improvement of Waste
Treatment Processes. Federal Water Pollution Control Administration, Grant
No. WPRD 102-01-68 (May, 1969).
6. Hazen and Sawyer, Engineers, Upgrading Existing Wastewater Treatment Facilities.
Prepared for U.S. EPA Technology Transfer Design Seminar, Pittsburgh, Pennsylvania
(August 29-31, 1971).
7. 1970 Annual Report, Livermore Water Reclamation Plant, City of Livermore,
California.
8. Brown and Caldwell Consulting Engineers, Project Report for Water Reclamation
Plant, Central Contra Costa Sanitary District, California (November, 1971).
9. West, A.W., Case Histories of Plant Improvement by Operations Control, Nutrient
Removal and Advanced Waste Treatment. Federal Water Pollution Control
Administration, Cincinnati, Ohio (1969).
10. Schwinn, D.E., Design Features of the District of Columbia's Water Pollution Control
Plant. Presented at the Sanitary Engineering Specialty Conference, ASCE, Sanitary
Engineering Division, Rochester, New York (June, 1972).
13-37
-------
-------
APPENDIX A
METRIC CONVERSION CHART
Multiply BX To Get
Inches 2.54 Centimeters
Feet 0.3048 Meters
Square Feet 0.0929 Square Meters
Cubic Feet 0.0283 Cubic Meters
Pounds 0.454 Kilograms
Gallons 3.79 Liters
Gallons/Minute 5.458 Cubic Meters/Day
Feet/Second 0.305 Meters/Second
A-l
-------
-------
APPENDIX B
WORD ABBREVIATIONS
Text
Tables
afternoon
average
before noon
biochemical oxygen demand
brake horsepower
centimeter(s)
chemical oxygen demand
cubic foot
cubic feet per minute
cubic feet per second
degree Centigrade
degree Fahrenheit
dissolved oxygen
elevation
feet per minute
food to microorganism ratio
foot (feet)
gallon(s)
gallons per day
gallons per minute
gram(s)
head loss
horsepower
hour(s)
inch(es)
Jackson turbidity units
kilowatt-hour
micron(s)
milligrams per liter
milliliter
millimeter(s)
million gallons
million gallons per day
PM
AM
BOD
BHP
cm
COD
cuft
cfm
cfs
degC
degF
DO
El
fpm
F/M
gpd
gpm
::
hp
kwh
y
mg/1
ml
mm
mil gal
PM
avg
AM
BOD
BHP
cm
COD
cuft
cfm
cfs
°C
DO
El
fpm
F/M
ft
gal
gPd
gpm
g
H.L.
hp
hr
in
Jtu
kwh
y
mg/1
ml
mm
mil gal
mgd
B-l
-------
APPENDIX B - Continued
Text Tables
mixed liquor suspended solids MLSS MLSS
mixed liquor volatile suspended solids MLVSS MLVSS
most probable number MPN
number (s) No.
parts per million PPm PPm
pound(s) (weight) Ib
pounds per cubic foot pcf pcf
pounds per square foot psf psf
pounds per square inch gage psig psig
revolutions per minute rpm rpm
second(s) sec
side water depth SWD
sludge retention time SET SRT
square foot sq ft sq ft
standard oxygen transfer rate SOR SOR
suspended solids SS SS
total solids TS
total suspended solids TSS
volatile suspended solids VSS
year(s) yr
B-2
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