vvEPA
United States
Environmental Protection
Agency
Office of Research and
Development
Washington DC 20460
EPA/600/R-95/051
April 1995
RCRA Subtitle D (258)
Seismic Design
Guidance for Municipal
Solid Waste Landfill
Facilities
-------
EPA/600/R-95/051
April 1995
RCRA SUBTITLE D (258)
SEISMIC DESIGN GUIDANCE
FOR
MUNICIPAL SOLID WASTE LANDFILL FACILITIES
by
Gregory N. Richardson
G.N. Richardson & Associates
Raleigh, North Carolina 27603
and
Edward Kavazanjian, Jr.
and
Neven Matasovi
GeoSyntec Consultants
Huntington Beach, California 92647
Contract No. 68-C3-0315
PROJECT MANAGER
Robert Landreth
Waste Minimization, Destruction and
Disposal Research Division
Risk Reduction Engineering Laboratory
Cincinnati, Ohio 45268
RISK REDUCTION ENGINEERING LABORATORY
OFFICE OF RESEARCH AND DEVELOPMENT
U.S. ENVIRONMENTAL PROTECTION AGENCY
CINCINNATI, OHIO 45268
Printed on Recycled Paper
-------
DISCLAIMER
The information in this document has been funded wholly or in part by the United States
Environmental Protection Agency (EPA) under Contract No. 68-C3-0315 WA #05 to Harding
Lawson Associates. It has been subjected to the Agency's peer and administrative review, and
it has been approved for publication as an EPA document. Mention of trade names or commercial
products does not constitute endorsement or recommendation for use.
n
-------
FOREWORD
Today's rapidly developing technologies, industrial products, and practices frequently
carry with them generation of materials that, if improperly dealt with, may threaten both human
health and the environment. The United States Environmental Protection Agency (EPA) is
charged by Congress with protecting the Nation's land, air, and water resources. Under a
mandate of national environmental laws, the Agency strives to formulate and implement actions
leading to a compatible balance between human activities and the ability of natural resources to
support and nurture life. These laws direct the EPA to conduct research to define our
environmental problems, measure the impacts, and search for the solutions.
The Risk Reduction Engineering Laboratory is responsible for planning, implementing,
and managing research, development, and demonstration programs. These programs provide an
authoritative, defensible engineering basis in support of the policies, programs, and regulations
of the EPA with respect to drinking water, wastewater, pesticides, toxic substances, solid and
hazardous wastes, and Superfund-related activities. This publication presents information on
current research efforts and provides a vital communication link between the researcher and the
user community.
Recent RCRA Subtitle D regulations (40 CFR Part 258) establish the requirements that
MSW landfills must not be sited where they can be damaged by active ground faulting (258.13)
and that they must be designed to resist the effect of regional earthquakes (258.14). This
document is intended to provide technical guidance to regulatory reviewers and landfill designers
to ensure these objectives are accomplished. It is meant to be a practical design document
applicable to the vast majority of MSW landfills.
Further information relative to this document may be obtained by writing
Robert Landreth, Risk Reduction Engineering Laboratory, Cincinnati, OH, 45268.
E. Timothy Oppelt, Director
Risk Reduction Engineering Laboratory
in
-------
ABSTRACT
On October 9, 1993, the new RCRA Subtitle D regulations (40 CFR Part 258) went into
effect. These regulations are applicable to landfills receiving municipal solid waste (MSW) and
establish minimum Federal criteria for the siting, design, operation, and closure of MSW landfills.
These regulations apply to the entire waste containment system, including liners, leachate
collection systems, and surface water control systems. This document presents field and design
procedures to satisfy the earthquake (or seismic) related criteria contained within these
regulations. Sample analyses are provided to evaluate the Subtitle D seismic requirements for a
range of site and facility conditions.
Section 258.13 of the regulations requires that new or lateral expansions of existing
landfills cannot be sited within 200-feet of a fault that has been active during the Holocene Epoch
(past 11,000 years) unless it can be demonstrated that a lesser setback is safe. This document
presents field identification methods used to identify active faults. Additionally, the document
reviews general tectonic and seismological considerations that strongly suggest that movement of
faults during the Holocene Epoch is very rare east of the Rocky Mountains.
Section 258.14 of the regulations identifies seismic impact zones within the United States
based on earthquake probability maps prepared by the United States Geological Survey (USGS).
Seismic impact zones are defined in the new regulations as those regions having a peak bedrock
acceleration exceeding 0.1 g based on a 90% probability of non-exceedance over a 250 year time
period. Within seismic impact zones, the regulations require that the waste containment system
for new MSW landfills and for lateral expansions of existing MSW landfills be designed to resist
the maximum horizontal acceleration in lithified earth material (MHA). The MHA is defined as
the maximum expected horizontal acceleration either depicted on a seismic hazard map with a 90
percent probability of non-exceedence in 250 years or based upon a site-specific seismic risk
assessment.
This document presents analysis procedures to evaluate the ability of the site subgrade
to resist liquefaction and of the waste mass/subgrade to resist slope failure where subjected to the
MHA. Sample calculations are provided to demonstrate the analysis techniques for liquefaction
and slope stability. Additional discussion is provided regarding more sophisticated deformation
analysis methods that may be required for MSW landfills in highly seismic regions.
This report was submitted in fulfillment of Contract No. 68-C3-0315 WA #05 under the
sponsorship of the United States Environmental Protection Agency. This report covers a period
from November, 1993 to May, 1994, and work was completed as of May, 1994.
IV
-------
TABLE OF CONTENTS
Number Page No.
Disclaimer ii
Foreword iii
Abstract iv
List of Tables vii
List of Figures viii
Abbreviations and Acronyms xii
Acknowledgements xiii
1.0 Introduction 1
1.1 Introduction to Subtitle D Seismic Criteria 1
1.1.1 Part 258.13 Fault Zone Siting Criteria 1
1.1.2 Part 258.14 Seismic Impact Zones 2
1.2 Scope of this Document 2
1.3 Limitations of this Document 4
1.4 References 4
2.0 258.13 Fault Area Considerations 7
2.1 Regional Fault Characteristics 7
2.2 Site Fault Characterization 9
2.3 Defining Fault Movement in Holocene Epoch 11
2.4 Comments on Fault Considerations East of the Rockies 12
2.5 References 12
3.0 258.14 Seismic Impact Zones: U.S.G.S. Probabilistic
Bedrock Acceleration 25
3.1 Development of Design Earthquake 26
3.2 Interpretation of Peak Bedrock Accelerations 27
3.3 References 28
4.0 258.14 Seismic Impact Zones: Site Specific Seismic Design
Ground Motion 40
4.1 General Methodology 42
4.1.1 Simplified Analysis 43
4.1.2 One-Dimensional Site Response Analysis 47
-------
TABLE OF CONTENTS (continued)
Number Page No.
4.1.3 Two- and Three-Dimensional Site Response Analysis 51
4.2 Selection of Earthquake Time History 52
4.3 References 55
5.0 258.14 Seismic Impact Zones: Liquefaction Analysis 74
5.1 Initial Screening 74
5.2 Liquefaction Potential Assessment 76
5.3 Liquefaction Impact Assessment 80
5.4 Liquefaction Mitigation 81
5.5 References 82
6.0 258.14 Seismic Impact Zones: Slope Stability and Deformation Analysis .... 102
6.1 Key Material Properties 103
6.1.1 Unit Weight 103
6.1.2 Interface Shear Resistance 104
6.1.3 Low Permeability Soil 104
6.1.4 Granular Soil Shear Strength 105
6.1.5 MSW Shear Strength 105
6.1.6 Sensitivity Studies 106
6.2 Seismic Stability and Deformation Analysis 106
6.3 Additional Considerations 109
6.4 References 110
Appendices
A - Seismic Design Examples - Liquefaction 125
B - Seismic Design Examples - Slope Stability 132
VI
-------
LIST OF TABLES
Number Page No.
2.1 Significant Earthquakes in Eastern North America (Adams and Busham, 1994) .. 15
2.2 Sources of Information 16
2.3 Addresses of State Geological Survey Offices Source (Geotimes, 1980) 17
2.4 Detailed Seismic Event Data Available from USGS National Earthquake Information
Center 18
3.1 Parameters for Seismic Source Zones (USGS, 1982) 31
4.1 Parameters for the Empirical Relationship to Estimate Gmax (after Imai and
Tonouchi, 1982) 59
5.1 Estimated Susceptibility of Sedimentary Deposits to Liquefaction During Strong
Seismic Shaking (Youd and Perkins, 1978) 87
5.2 Recommended "Standardized" SPT Equipment (after Seed, et al., 1985 and Riggs,
1986) 88
5.3 Correction Factors for Nonstandard SPT Procedure and Equipment 89
5.4 Improvement Techniques for Liquefiable Soil Foundation Conditions (NRC,
1985) 90
6.1 Unit Weight Data for MSW (Fassett et al., 1994.) 114
6.2 Compilation of the Available Shear Strength Data on MSW (GeoSyntec, 1993) . 116
6.3 Lower Bound Friction Angles Backfigured from Observations of Steep Landfill
Slopes () (Kavazanjian et al., 1995) 117
vu
-------
LIST OF FIGURES
Number Page No.
1.1 Seismic Impact Zones (Areas with a 10% or Greater Probability that the Maximum
Horizontal Acceleration will Exceed 0.10 g in 250 Years) (EPA, 1993) 6
2.1 The Six Major Tectonic Plates and their Approximate Linear Velocity Vectors
(Adapted from Park, 1983) 19
2.2 Seismic Source Areas in the United States (Krinitsky, et al., 1993) 20
2.3 Isoseismal Contours for Intra-Plate vs. Edge-Plate Events of Similar Magnitude
(Nuttli, 1974) 21
2.4 Epicenters for Earthquakes M 2.5 in the Southeastern United States (July 1977 -
December 1984) (Sibol et al., 1984) 22
2.5 Characteristics of Hay ward Fault as Exposed in Five Trenches at Fremont Site
(Cluff et al., 1972) 23
2.6 Detail of West Fault Trace Exposed in Trench "G" at Fremont Site (see Fig. 2.5)
(Cluff et al., 1972) 24
3.1 Basic Elements of the USGS Probabilistic Hazard Calculations: (a) Typical Source
Areas and Grid of Points at which the Hazard is to be Computed; (b) Statistical
Analysis of Seismicity Data and Typical Attenuation Curves; (c) Cumulative
Conditional Probability Distribution of Acceleration; (d) The Extreme Probability,
Fmax.t (a) for Various Accelerations and Exposure Times (T) (USGS, 1982) .... 35
3.2 Seismic Source Zones in the Contiguous United States (USGS, 1982) 36
3.3 Seismic Source Zones in the Central United States (Johnson and Nava, 1994) ... 37
3.4 Time-Dependent Fluctuations in Seismic Ground Response Parameters (17 January
1994 Northridge, California Earthquake, Oil Site, Longitudinal Component)
(Hushmand Associates, 1994) 38
3.5 Contribution of Various Magnitudes and Distances to the Seismic Hazard
(Moriwaki et al., 1994) . 39
vni
-------
LIST OF FIGURES (continued)
Page No.
4. 1 Soil Conditions and Characteristics of Recorded Ground Motions, San Francisco
M 5.7 Earthquake of 22 March 1957 (Seed, 1968) ..................... 60
4.2 Development of Acceleration Response Spectrum for Damped Single Degree of
Freedom System .......................................... 61
4.3 Tripartite Representation of Response Spectra ........................ 62
4.4 Relationship Between Maximum Acceleration on Rock and Other Local Site
Conditions: (a) Seed and Idriss (1982); (b) Idriss (1990) ................. 63
4.5 Observed Variations of Peak Horizontal Accelerations on Soft Soil and MSW
Sites in Comparison to Rock Sites (Kavazanjian and Matasovi, 1995) ......... 64
4.6 Approximate Relationship Between Maximum Ground Accelerations at the Base and
Crest for Various Ground Conditions (Singh and Sun, 1995) .............. 65
4.7 Variation of Maximum Average Acceleration Ratio with Depth of Sliding Mass
(Kavazanjian and Matasovi, 1995) ............................... 66
4.8 Normalized Maximum Horizontal Equivalent Acceleration versus the Normalized
Fundamental Period of the Waste Fill (Bray et al., 1995) ................ 67
4.9 Modulus Reduction and Damping Curves for Soils of Different Plasticity Index (PI)
(Vucetic and Dobry, 1991) .................................... 68
4.10 Shear Wave Velocity of MSW (Kavazanjian et al., 1995) ................ 69
4.11 Modulus Reduction and Damping Curves for MSW (Earth Technology, 1988) ... 70
4.12 Modulus Reduction and Damping Curves for MSW (Singh and Murphy, 1990) ..71
4.13 Modulus Reduction and Damping Curves for MSW (Kavazanjian and Matasovi,
1995) .................................................. 72
4. 14 Comparison of OH Landfill Response to Results of Equivalent Linear Analysis
(Kavazanjian et al., 1995) ..................................... 73
IX
-------
LIST OF FIGURES (continued)
Number Page No.
5.1 Grain Size Distribution Curves of Liquefied Soils (Ishihara et al., 1989) 92
5.2 Variation of q^N^ Ratio with Mean Grain Size, D50 (Seed and De Alba, 1986) . . 93
5.3 Stress Reduction Factor, rd (Seed and Idriss, 1982) 94
5.4 Correction Factor for the Effective Overburden Pressure, CN (Seed etal., 1983) . 95
5.5 Relationships Between Stress Ratio Causing Liquefaction and (N,)^ Values for Sand
for M 7.5 Earthquakes (Seed et al., 1985) 96
5.6 Curve for Estimation of Magnitude Correction Factor, kM (after Seed et al., 1983) 97
5.7 Curves for Estimation of Correction Factor k (Harder 1988, and Hynes 1988, as
Quoted in Marcuson et al., 1990) 98
5.8 Curves for Estimation of Correction Factor k (Harder 1988, and Hynes 1988, as
Quoted in Marcuson et al., 1990) 99
5.9 Curves for Estimation of Post-Liquefaction Volumetric Strain using SPT Data and
Cyclic Stress Ratio (Tokimatsu and Seed, 1987) 100
5.10 Relationship Between Corrected "Clean Sand" Blowcount (N,)^ and Undrained
Residual Strength (Sr) from Case Studies (Seed et al., 1988) 101
6.1 Fundamental Principles of the Newmark Seismic Deformation Analysis (after Bray et
1994) 118
6.2 Unit Weight Profile for MSW (Kavazanjian et al., 1995) 119
6.3 Bi-Linear Shear Strength Envelope for MSW (Kavazanjian etal., 1995) 120
6.4 Yield Acceleration as a Function of Shear Strength Parameters for the Oil Landfill
(Siegel et al., 1990) 121
6.5 Hynes and Franklin Permanent Seismic Displacement Chart (Hynes and Franklin,
1984) 122
-------
LIST OF FIGURES (continued)
Number Page No.
6.6 Makdisi and Seed Permanent Displacement Chart (Makdisi and Seed, 1978) . . . 123
6.7 Modes of Instability of a MSW Landfill 124
XI
-------
ABBREVIATIONS AND ACRONYMS
ASTM American Society for Testing and Materials
ATC Applied Technology Council
CPT Cone Penetration Test
CSR Critical Stress Ratio
EERC Earthquake Engineering Research Center
EERI Earthquake Engineering Research Institute
EPA United States Environmental Protection Agency
FSAR Final Safety Analysis Report
MFZ Mendocino Fracture Zone
MHA Maximum Horizontal Acceleration
MM Modified Mercali (Intensity Scale)
MSW Municipal Solid Waste
MSWLF Municipal Solid Waste Landfill Facility
NAPP National Aerial Photographic Program
NCEER National Center for Earthquake Engineering Research
NRC National Research Council
Oil Operating Industries, Inc. (Landfill)
PSAR Preliminary Safety Analysis Report
RCRA Resource Conservation and Recovery Act
SDOF Single Degree of Freedom (System)
SPT Standard Penetration Test
SSA Seismological Society of America
USGS United States Geological Survey
xn
-------
ACKNOWLEDGEMENTS
The authors wish to express their sincere appreciation to the following individuals who reviewed
and critiqued drafts of this manuscript: Dr. Rudolph Bonaparte, GeoSyntec Consultants,
Professor Jonathan Bray, University of California of Berkeley, Professor Ron Chancy, Californis
State University at Humbolt, Professor Robert Koerner, Drexel University, Professor Gerald
Leonards, Purdue University, Mr. Robert Phaneuf, New York State Department of Environmental
Conservation, and Mr. Robert Landreth, United States Environmental Protection Agency. The
authors also thank the New York Department of Environmental Conservation for providing case
studies used in the sample calculations.
The authors also gratefully acknowledge the many individuals, too numerous to name here, who
over the years have shared their experiences and recommendations regarding seismic probability
studies, liquefaction analysis, and dynamic stability evaluation.
Xlll
-------
SECTION 1
INTRODUCTION
On October 9, 1993, the RCRA Subtitle D regulations (40 CFR Part 258) went into effect. These
regulations are applicable to landfills receiving municipal solid waste (MSW) and establish
minimum Federal criteria for the site location, design, operation, ground-water monitoring, and
closure/post closure care of MSW landfills. This document focuses on the earthquake (or seismic)
siting and facility design criteria contained within Subtitle D. The document is intended for use
by both designers of MSW landfills and the regulatory community that reviews such designs.
Where possible, actual landfill situations have been used in the development of example problems
to demonstrate the various analysis procedures. Emphasis is placed herein on simple analysis
methods that are within the technical capabilities of the general engineering community. The
range of applicability and the limitations of these methods are reviewed and more rigorous
analysis methods are briefly summarized.
1.1 Introduction to Subtitle D Seismic Criteria
Subtitle D regulations address the potential for damage to a MSW landfill resulting from relative
ground displacements (e.g., fault displacement) and from strong ground motions (e.g., ground
accelerations) that can accompany an earthquake. Limiting the potential for fault displacement-
induced damage is accomplished by siting criteria (258.13) that may preclude the use of a given
site for a MSW landfill. The impact of earthquake-induced strong ground motions on a MSW
landfill must be addressed by the design engineer. Subtitle D does not specify the required
evaluation process but establishes (258.14) a lower value for the maximum horizontal acceleration
(MHA) in lithified earth material (e.g. the peak bedrock acceleration) that must be considered in
the design of landfill containment structures. The MHA may be based upon either a probabilistic
map such as those published by the United States Geological Survey (USGS) or upon the results
of a site-specific analysis. Landfill containment structures are defined to include liners, leachate
collection systems, and surfaces water control systems.
1.1.1 Part 258.13 Fault Zone Siting Criteria
The Federal Subtitle D regulations state that a new MSW landfill or a lateral expansion of an
existing landfill may not be located within 200 feet (60 meters) of a fault that has experienced
displacement in the Holocene time unless the owner or operator demonstrates to the Director of
an approved State Program that an alternative setback distance of less that 200 feet (60 meters)
will prevent damage to the structural integrity of the landfill unit and will be protective of human
health and the environment. Within the regulations, a fault means a fracture or zone of fractures
1
-------
along which strata from one side have been displaced with respect to strata on the other side. The
Holocene time means the most recent epoch of the Quaternary period, e.g. within the last 10,000
to 12,000 years. This requirement means that MSW landfill site suitability studies must both
identify potential fault zones that impact the proposed site and then evaluate whether fault
displacement has occurred during the past 10,000 to 12,000 years. Section 2.0 of the document
presents the technical methodology for identifying fault zones and for complying with the
regulatory criteria.
1.1.2 Part 258.14 Seismic Impact Zones
A seismic impact zone is defined in the Subtitle D regulations as an area having a 10% or greater
probability that the peak horizontal acceleration in lithified earth material, expressed as a
percentage of the earth's gravitational pull (g), will exceed 0.10 g in 250 years. These zones may
be defined using seismic probability maps prepared by the USGS (USGS, 1982 and USGS, 1991)
or by more detailed regional or site specific studies. The USGS maps present peak bedrock
accelerations and velocities reflecting a 90% probability that the acceleration will not be exceeded
over 10, 50 and 250 year interval periods. Seismic impact zones in the United States, defined by
application of the Subtitle D criteria to the USGS seismic probability maps, are shown in
Figure 1.1.
Section 3.0 of this document provides general background information on the development of the
USGS seismic probability maps, and a simple method for interpretation and use of the peak
bedrock acceleration from these maps. Section 4.0 provides methodologies for calculating the
peak ground surface acceleration at a landfill site and the peak surface acceleration and peak
average acceleration of the waste mass based on site characteristics and peak bedrock acceleration.
These peak ground accelerations are then used in Section 5.0 for evaluating the liquefaction
potential of a site and in Section 6.0 for evaluating the stability of a landfill foundation, waste
mass, and waste slopes. Sections 5.0 and 6.0 present simplified seismic analysis procedures that
can typically be performed without the need for supplemental field investigative programs,
expensive specialized laboratory testing, or sophisticated dynamic analyses to evaluate compliance
of the design of MSW landfills with Subtitle D regulation for seismic design.
1.2 Scope of This Document
Damage to landfills from earthquake may be due to the primary seismic hazard of fault
displacement or to secondary hazards such as slope instability or liquefaction of the foundation
induced by strong ground motions. Potential modes of damage MSW landfills associated with the
primary seismic hazard include:
-------
• disruption of liner and cover systems;
• disruption of the landfill gas control system; and
• disruption of surface water and drainage control systems.
Secondary modes of damage to the containment systems of MSW landfills that are subject to
strong ground motions include:
• damage due to liquefaction and lateral spreading of the foundation;
• damage due to seismically induced settlement of the foundation; and
• damage due to seismically-induced landslides.
In general, MSW landfills have performed extremely well in earthquakes. Observations of the
performance of solid waste landfills subject to strong ground motions (Anderson and Kavazanjian,
1995; Matasovic, et al., 1995) indicate that minor cracking of cover soils at the waste/natural
ground interface and disruption of landfill gas control systems due to loss of power and breaking
of vertical wells and headers are the most common types of damage experienced by MSW landfills
subject to strong ground shaking. Neither of these effects is considered to present a significant
environmental hazard. However, experience with the performance of modern landfills conforming
to Subtitle D requirements is limited. Of the three landfills designed in accordance with Subtitle
D standards subject to the strongest shaking in the Northridge earthquake of 17 January 1994, one
experienced two tears in the liner, one of which was approximately 75 ft (23 m) in length, along
an anchor trench above the waste. Furthermore, no landfill with a geosynthetic cover is known
to have been subjected to strong shaking in an earthquake and no solid waste landfill is known to
have experienced fault displacement or liquefaction in the foundation during an earthquake (even
though there are solid waste landfills known to be sited on active faults and liquefiable soils).
Therefore, caution is warranted in concluding unconditionally that landfills will continue to
perform well in earthquakes and investigations and analyses are required to demonstrate that
landfills are properly sited to avoid active faults and are properly designed to resist the effects of
strong ground motions and liquefaction
This document presents a set of simplified analyses for seismic performance analysis of the waste
mass, liner and cover systems, and foundation of a MSW landfill within a seismic impact zone.
The analyses presented herein include analyses of the impact of instability of the waste mass and
cover soil on the integrity of geosynthetic liner and cover systems of a landfill subject to strong
ground motions. Analyses of the potential for liquefaction-induced lateral spreading of the waste
mass and seismically-induced settlement of the foundation soil are also presented herein. The
simplified analyses presented herein provide a minimum standard for design of MSWLF in
accordance with Subtitle D standards. These analyses are not intended for analysis of the seismic
performance of landfills containing hazardous or toxic substances or large amounts of liquids.
-------
Neither should they be applied to landfills that do not conform to Subtitle D siting restrictions.
Such landfills must be considered on a case-by-case basis and may require a higher standard of
care dependency on the potential consequences of seismically-induced damage.
1.3 Limitations of this Document
The simplified analyses described in this document are presented as an example of one way in
which such a seismic performance assessment may be conducted. The simplified analyses
presented herein are designed to produce an expedient assessment of the seismic resistance of the
landfill containment systems. If such simplified analyses indicate potential seismic problems
(e.g., results in unacceptable factors of safety), then more sophisticated analysis methods may be
required to demonstrate satisfactory performance of the facility.
This document addresses the seismic design of the landfill waste mass, liner and cover systems,
and foundations, only. With the exception of the liquefaction analyses, this document does not
provide guidance on assessing the impact of geologically unstable terrain on landfill performance.
A demonstration that the MSW landfill unit will not be disrupted by geologic instability, including
seismically unstable areas, is required under Section 258.15 of Subtitle D. Guidance for satisfying
the provisions of Section 258.15 of Subtitle D is provided by EPA (1993) and is not included in
this document. Additional seismic analyses may be required to assess the performance of other
components of the landfill containment systems, including the leachate collection system and
surface-water control systems.
Seismic analysis and design of landfills is a rapidly developing field. Even as this document was
being completed, new and important studies on the seismic behavior of landfills were appearing
in print and/or being presented at conferences and other professional meetings. It is essential that
the designer and regulator involved in seismic design of MSW landfills keep abreast of current
developments in the field. As new information and techniques become available, they will
supersede the information and methods presented herein.
1.4 References
Anderson, D.G., and Kavazanjian, E. Jr. (1995) "Performance of Landfills Under Seismic
Loading," Proc., Third International Conference on Recent Advances in Geotechnical Earthquake
Engineering and Soil Dynamics, University of Missouri, Rolla, Vol. 3, 2-7 April.
EPA (1993), "Technical Manual: Solid Waste Disposal Facility Criteria," United States
Environmental Protection Agency, EPA 530-R93-017, Washington, District of Columbia.
-------
Matasovic, N., Kavazanjian, E., Jr., Augello, A.J., Bray, J.D., and Seed, R.B. (1995), "Solid
Waste Landfill Damage Caused by 17 January 1994 Northridge Earthquake," In: Woods, Mary C.
and Seiple, Ray W., Eds., The Northridge, California, Earthquake of 17 January 1994, California
Department of Conservation, Division of Mines and Geology Special Publication 116,
Sacramento, California, pp. 43-51.
USGS (1982), "Probabilistic Estimates of Maximum Acceleration and Velocity in Rock in the
Continuous United States," United States Geological Survey, Open-File Report 82-1033.
USGS (1990), "Probabilistic Earthquake Acceleration and Velocity Maps for the United States
and Puerto Rico," United States Geological Survey, Miscellaneous Field Studies Map MF-2120.
-------
Figure 1.1 Seismic Impact Zones (Areas With a 10% or Greater Probability that the
Maximum Horizontal Acceleration Will Exceed 0.10 g in 250 Years) (EPA,
1993).
-------
SECTION 2
258.13 FAULT AREA CONSIDERATIONS
Locating a landfill in the vicinity of faults that have experienced relative movement in recent times
poses significant risk to the integrity of the landfill containment system. The impact to the landfill
from a seismic event can result directly from ground surface rupture or from deformation,
liquefaction, lateral spreading, and differential settlement induced by ground shaking that
accompanying the event. The fault area location restrictions imposed by Subtitle D restrict siting
of new MSW units or lateral expansions of existing units within 200 feet (60 meters) of a fault that
has displaced in Holocene time.
This section of the guidance document reviews methods for evaluating both the presence of faults
on-site and the possible movement of a fault within the Holocene Epoch. The section concludes
with a discussion regarding the difficulty in applying such methodology to faults located east of
the Rocky Mountains.
2.1 Regional Fault Characteristics
Faults are created when the stresses within geologic materials exceed the ability of those materials
to withstand the stresses. An understanding of such stresses is aided by a review of current plate
tectonics theory. Figure 2.1 shows the major tectonic plates that form the earth's continents and
their directions of movement. Along the west coast, earthquakes are the result of several different
fault systems that occur along the edge of the Pacific and North American plates. South of the
Mendocino Fracture Zone (MFZ) approximately 200 miles (320 kilometers) north of San
Francisco, the San Andreas fault system (strike-slip) controls earthquakes. North of the MFZ,
earthquakes are controlled by the Cascadia Subduction Zone. In between these two major fault
systems lie the ridges, rifts, and subduction zones associated with the Juan de Fuca and Gorda
plates. The complex interactions between these zones create stresses in the crust away from the
plates that generate earthquakes. Earthquakes may occur along faults in the crust in Washington,
Oregon, and California adjacent to the plate boundaries or in the interior of the continent away
from the plate boundaries. In the interior of the North American plate, tectonic stresses have
created fault systems that are known to have generated major earthquakes in Utah, South Carolina,
New England, Oklahoma, and Missouri/Tennessee in Holocene time. The sense of the fault
displacement within these fault systems range from horizontal (strike-slip) to vertical (dip-slip) to
combinations of these components. These major fault systems and one suggested representation
of the major seismic source areas in the United States are shown in Figure 2.2. The Roman
numerals on this figure represent the maximum observed (historic) seismic intensity in the region
as measured by the Modified Mercali (MM) intensity scale (Richter, 1958).
-------
In contrast to the west coast, earthquakes east of the Rocky Mountains cannot be associated with
the relative displacements of edge-plate faults (active margin). Intra-plate (passive margin)
earthquakes occur less frequently than the edge-plate associated earthquakes of the west coast but
impact a significantly larger geographic area. Table 2.1 (Adams and Busham, 1994) presents one
summary of significant earthquakes in Eastern North America in historic time, ordered by
decreasing magnitude. As this table shows, the pattern of significant earthquakes east of the
Rocky Mountains is broadly dispersed both geographically and temporally.
The differences hi the sizes of affected areas may be caused by the differences in stress conditions
in the basement rock structure. In the west, the stress condition is predominantly tension, while
in the east, stresses hi the basement rock are primarily compressional. Whatever the mechanism,
the rate of attenuation of earthquake ground motions east of the Rocky Mountains appears to be
significantly slower than in the western United States (Nuttli, 1974; 1981), resulting in a much
larger impacted area in the eastern U.S. than the western U.S. for earthquakes of the same
magnitude. For equivalent historical earthquakes, Figure 2.3 shows the isoseismal contours of
MM VI and VII for an event at the plate boundary in the western United States and for two intra-
plate events in the eastern United States. Note the small geographic area impacted by the western
edge-plate event as compared to the two intra-plate event; one that shook a large portion of the
central United States centered around New Madrid and one that shook far beyond Charleston,
even into Canada. Another observed difference between earthquakes in the eastern and western
U.S. is that eastern earthquakes appear to be enriched in high frequency components compared
to western earthquakes (Atkinson, 1987).
The significance of the differences between western edge-plate earthquakes and the intra-plate
events that occur east of the Rockies with respect to identification of surface faulting is discussed
subsequently within this section. The significance of the differences between western and eastern
earthquakes with respect to frequency content is discussed in subsequent sections.
Characterization of the seismicity of the eastern and central U.S. is a topic of much current study
and discussion (Applied Technology Council (ATC), 1994). Due to the many ongoing studies,
our understanding of the seismicity of the central and eastern United States is evolving rapidly.
Prudent investigators should consult current sources of information on local and regional
seismicity at the initiation of any project. Sources of current information are discussed
subsequently in this section.
-------
2.2 Site Fault Characterization
The principal factors controlling the general characteristics of surface faulting are: (a) the type of
fault (reverse, normal, or strike-slip), (b) the inclination of the fault plane, (c) the amount of
displacement on the fault, (d) the depth and geometry of the surficial earth materials, and (e) the
nature of the overlying earth materials. Strike-slip faults that are not fairly linear may produce
complex surface features. Step-over zones where fault displacement is transferred from adjacent
strike-slip faults may be particularly complex. Dip-slip faults, with either normal or reverse
motion, typically produce multiple fractures within rather wide and irregular fault zones. These
zones generally are confined to the hanging-wall side of the fault leaving the footwall side little
disturbed. With respect to fault impacts on a structure, setback requirements for such faults may
be rather narrow on the footwall side, depending on the quality of data available, and larger on
the hanging wall side of the zone. Some fault zones may contain broad deformational features
such as pressure ridges and sags rather than clearly defined fault scarps or shear zones (Hart,
1990).
An investigation to identify faulting at a given site must rely on a review of available data and
field geologic reconnaissance methods. Available data may include pertinent technical
publications, unpublished reports, maps, aerial photographs, and interviews with experts familiar
with the region under study. Pertinent technical publications include maps prepared by the USGS
identifying young faults in the western states, publications of the Seismological Society of
America, and regional reports from the seismological networks and state geological surveys. A
detailed summary of available sources of engineering geologic information is presented by
Trautmann and Kulhawy (1983). General sources for such information are indicated on
Table 2.2. Table 2.3 provides a listing of addresses of the geological survey offices for all
50 states.
Studies performed for siting of nuclear power plants can be a useful source of information on
regional seismicity and geology. All applications for construction permits for nuclear generating
stations are required to submit documentation on regional geology, including known faults and
observed seismicity, within a 200 mile (320 kilometer) radius of the site. This information can
be found in the Preliminary Safety Analysis Report (PSAR) and the Final Safety Analysis Report
(FSAR) for the project. These reports are available through the National Technical Information
Service (see Table 2.2) for all existing and many proposed nuclear generating stations. However,
as may of these reports are over 20 years old, more recent sources of information on regional
seismicity and tectonics should be consulted.
Existing seismic networks provide very detailed identification of recent earthquakes within seismic
impact regions. Such information includes the magnitude and epicentral location of all identified
-------
events and is commonly available plotted in map form as shown on Figure 2.4. A detailed
evaluation of each detected event is also available as shown on Table 2.4. Note that while the
presence of micro-seismic activity can be used to infer the location of a subsurface fault, it cannot
be directly interpreted as evidence that surface displacement of the fault has taken or will take
place. To date, the only known earthquake east of the Rocky Mountains in historic time that has
been accompanied by observations of surface fault rupture is the 1989 Ungava, Quebec earthquake
of magnitude 6.3. The Meers fault in Oklahoma, where evidence points to a magnitude 7 +
earthquake within the past 1,100 to 1,400 years, and the Reelfoot fault in Tennessee, the source
of the 1811/1812 New Madrid earthquake sequence, are the only other generally recognized
Holocene faults east of the Rocky Mountains.
An interpretation of available stereo aerial photographs is useful in identifying and locating
potentially active faults. One source of such photographs is provided in Table 2.3. Other sources
are discussed by Trautmann and Kulhawy (1983). Active faults may be indicated in aerial
photographs by geomorphic features such as fault scarps, triangular facets, fault scarplets, fault
rifts, fault slice ridges, shutter ridges, and fault saddles (Cluff, et al. , 1972). Additional evidence
can be provided by ground features such a open fissures, offsets in fence lines, landscape features,
mole tracks and furrows, etc., rejuvenated streams, folding or warping of young deposits, ramps,
ground water barriers in recent alluvium, echelon faults in alluvium, and fault paths on young
surfaces. Usually a combination of such features is generated by recent fault movements at the
surface. Note that many of the fault movement indicators require the presence of undisturbed
surface soils at the site. Regions that have limited surface soils due to past geologic mechanisms
or man's activities can provide a significant challenge in demonstrating the recent activity of
existing faults. The aerial photo analysis should include an area within a five-mile radius of the
site.
Initial field reconnaissance should be performed at a minimum for the area within approximately
1-kilometer (3300-feet) of the proposed unit (EPA, 1993). This initial field reconnaissance can
include the following:
• walking portions of the site within 1-kilometer (3300 feet) of the unit to identify
possible geomorphic or ground features that indicate faulting;
• preparation and interpretation of special aerial photographs such as low sun angle
photographs that use shadows to accentuate topographic differences, infrared
photos that indicate differences in surface moisture content, and color photographs
to study slight color changes.
Section 2.3 discusses the field methods for establishing fault movement.
10
-------
2.3 Defining Fault Movement in Holocene Epoch
If the site fault characterization study indicates the potential presence of faults on the proposed
landfill site, then a detailed geologic reconnaissance may be required for the site of the proposed
unit. A detailed geologic surface reconnaissance should be made to identify the best
approximation of the fault location on site and the amount and sense of past fault movements.
The detailed field site characterization can include the following:
• using geophysical methods such as resistivity, seismic refraction, or magnetic
methods to identify specific fault locations;
• excavation of exploratory trenches at an angle to faults identified on the site to
allow the detailed examination of the trench walls for evidence of recent fault
displacements; and
• subsurface drilling exploration to locate fault zones.
The depth of the subgrade investigated should be sufficient to represent activities within the
Holocene Epoch. Radiocarbon dating of carbonaceous material encountered can be used to
constrain the age of most recent fault offsets. A detailed description of soil-stratigraphic dating
techniques is presented by Shlemon (1985). Sieh et. al (1984) describe the application of high-
precision radiocarbon analyses for chronological analysis of active faulting. Note that establishing
that recent displacement has occurred is greatly complicated if a limited soil profile over rock
exists at the site, e.g., glacially polished regions, or if the Holocene zone of the alluvium is absent
or disturbed.
Trenching across a fault through overlying alluvium and colluvium has been the most common
tool used to establish both the existence of fault displacement and for dating the displacement.
Trench geologic sections established by trenching for portions of the Hay ward fault in California
are shown on Figure 2.5. These trench sections established that the western trace of the Hay ward
fault was active (e.g., the fault displacements projected up through the overlying alluvium) and
that the eastern trace was not active. These observations are shown on Figure 2.6. Note the
distinct stratigraphic displacements that identify the west fault trace. Thus, this study would lead
to the requirement that all proposed MSW landfills be located more than 60 meters (200 feet) from
the western trace of the Hay ward fault. Since the eastern trace of the Hay ward fault was found
not to be active, there are no regulatory constraints that would exclude siting a MSW landfill
adjacent to the eastern trace of the Hayward fault on the basis of faulting.
11
-------
ATC (1994), "Seminar on New Developments in Earthquake Ground Motion Estimation and
Implications for Engineering Design Practice," Applied Technology Council, ATC35-1, Redwood
City, California.
Atkinson, G.M. (1987), "Implications of Eastern Ground Motion Characteristics for Seismic
Hazard Assessment in Eastern North America," Proc. Symposium on Seismic Hazards, Ground
Motions, Soil-Liquefaction and Engineering Practice in Eastern North America, Tuxedo, New
York, NCEER Technical Report No. NCEER-87-0025.
Cluff, L.S, Hansen, W.R., Taylor, C.L., Weaver, K.D., Brogan, G.E., Idriss, I.M., McClure,
F.E., and Blayney, J.A. (1972), "Site Evaluation in Seismically Active Regions - An
Interdisciplinary Team Approach," Proc. International Conference on Microzonation for Safety
Construction, Research and Application, Seattle, Washington, Vol. 2, p. 9-57 - 9-87.
Engdahl, E.R., and Rinehard, W.A. (1988), "Seismicity Map of North America," Geologic
Society of America, Centennial Special Map CSM-4, Scale 1:5,000,000.
EPA (1993), "Technical Manual: Solid Waste Disposal Facility Criteria" United States
Environmental Protection Agency, EPA 530-R93-017, Washington, District of Columbia.
Geotimes (1980), "A Directory of Societies in Earth Science," Vol. 25, No. 7, pp. 21-29.
Hart, E.W. (1990), "Fault-Rupture Zones in California," Special Publication 42, California
Division of Mines and Geology, Sacramento, California.
Himes, L., Stauder, W., and Herrmann, R.B. (1988), "Indication of Active Faults in the New
Madrid Seismic Zone from Precise Location of Hypocenters," National Workshop on
Seismogenesis in the Eastern United States, A.C. Johnston, et al, Eds. (also presented in
Seismological Research Letters, Vol. 59, No. 4, Seismological Society of America).
Johnston, A.C., and Nava, S.J. (1994), "Seismic Hazard Assessment in the Central United
States," Proc. Seminar on New Developments in Earthquake Ground Motion Estimation and
Implications for Engineering Design Practice, Applied Technology Council, ATC35-1, Redwood
City, California, pp. 2-1 - 2-12.
Krinitzsky, E.L., Gould, J.P. and Edinger, P.H. (1993), "Fundamentals of Earthquake-Resistant
Construction," John Wiley & Sons, New York, New York.
13
-------
Nuttli, O.W. (1974), "Seismic Hazard of the Rocky Mountains," Preprint 2/95, American Society
of Civil Engineers, National Structural Engineering Meeting, Cincinnati, Ohio.
Nuttli, O.W. (1981), "Similarities and Differences Between Western and Eastern United States
Earthquakes, and their Consequences for Earthquake Engineering," Earthquakes and Earthquake
Engineering: The Eastern United States, Vol. 1, Assessing the Hazard - Evaluating the Risk, J.E.
Beavers, Ed., Ann Arbor Science Publishers, Inc., Ann Arbor, Michigan, pp. 25-51.
Park, R.G. (1983), Foundations of Structural Geology, Blackie Publishing, Chapman and Hall,
New York, New York.
Richter, C.F. (1958), Elementary Seismology, W.H. Freeman and Company, San Francisco,
California.
Shlemon, R.J. (1985), "Application of Soil-Stratigraphic Techniques to Engineering Geology,"
Bulletin of the Association of Engineering Geologists, Vol. XXII, No. 2, pp. 129-142.
Sibol, M.S., Bellinger, G.A., and Mathena, E.G. (1984), "Seismicity of the Southern United
States," Southeastern U.S. Seismic Network Bulletin No. 15, Seismologic Observatory, Virginia
Polytechnic Institute and State University, Department of Geological Sciences, Blacksburg,
Virginia.
Sieh, K., Stuiver, M., and Brillinger, D. (1989), "A More Precise Chronology of Earthquakes
Produced by the San Andreas Fault in Southern California," Journal of Geophysical Research,
Vol. 94, No. Bl, pp. 603-623.
SSA (1988), "National Workshop on Seismogenesis in the Eastern United States," A.C. Johnston,
A.C. et al., Eds. (presented in Seismological Research Letters, Vol. 59, No. 4, Seismological
Society of America).
Trautmann, C.H., and Kulhawy, F.H. (1983), "Data Sources for Engineering Geologic Studies,"
Bulletin of the Association of Engineering Geologists, Vol. XX, No. 4, pp. 439-454.
14
-------
TABLE 2.1: SIGNIFICANT EARTHQUAKES IN
EASTERN NORTH AMERICA
SOURCE: ADAMS AND BUSHAM (1994)
Earthquake
New Madrid Region
New Madrid Region
New Madrid Region
Baffin Bay
Grand Banks
Charlevoix, Que
Charleston, SC
Nahanni, N.W.T.
Charlevoix, Que
Ungava, Que
Charleston, MO
Tuniskaming, Que
Charlevoix, Que
Cape Ann, offshore
Charlevoix, Que
New Madrid Region
Charlevoix, Que
Franklin L., N.W.T.
Saguenay, Quebec
Giles County, VA
Massena/Cornwall
Miramichi, N.B.
Attica, NY
Year
1812
1811
1812
1933
1929
1663
1886
1985
1870
1989
1895
1935
1925
1755
1791
1843
1860
1992
1988
1897
1944
1982
1929
M
8.7
8.6
8.4
7.3
7.2
7.0
6.9
6.9
6.5
6.3
6.2
6.2
6.2
6.1
6.0
6.0
6.0
6.0
5.9
5.8
5.8
5.7
5.5
Comments
largest stable craton eq.
largest Arctic earthquake
27 dead from tsunami
earliest large earthquake
devastating
prior M 6.6 event
10 km surface rupture
Quebec/Ontario border
might be larger
snaking equivalent to M 6.f
Ontario/NY border
shallow, three M 5 aftershock;
western NY
15
-------
TABLE 2.2: SOURCES OF INFORMATION
Aerial Photographs
Young Fault Maps
National Seismicity
National Technical
Information Service
National Aerial Photographic Program (NAPP)
National High Altitude Program (NHAP)
USGS EROS Data Center
Sioux Falls, South Dakota
(605) 594-6151
United States Geological Survey (USGS)
USGS National Center
H800) USA-MAPS or
USGS Map Sales Center
(303) 236-7477
National Earthquake Information Center
United States Geological Survey (USGS)
P.O. Box 25046, Denver Federal Center, MS 967
Denver, CO 80225
(800) 525-7848
Earthquake Engineering Research Institute (EERI)
499 14th Street, Suite 320
Oakland, California 94612
(510) 451-0905
National Center for Earthquake Engineering Research (NCEER)
State University of New York at Buffalo
Red Jacket Quadrangle
Buffalo, New York 14260
(716) 645-3391
Seismological Society of America (SSA)
National and Eastern Section
El Cerrito, California
(415) 525-5474
5285 Port Royal Road
Springfield, VA 22161
(703) 487-4650
FTS 737-4650
16
-------
TABLE 2.4: TYPICAL SEISMIC EVENT DATA AVAILABLE FROM
USGS NATIONAL EARTHQUAKE INFORMATION CENTER
AUGUST 19t*
« OAT
C
41
41
41
41
41
4,
41
41
41
41
41
41
41
41
41
«,
41
91
41
41
41
41
41
41
41
41
41
41
41
41
O*t«l« TIUC
UTC
M MN SCC
44 14 99. «
41
41
41
41
42
42
42
42
43
43
44
44
44
49
,4
44
44
44
••
M
44
(1
••
••
,»
14
14
11
11
12
14
14
a*
43
12
44
94
44
47
3*
1*
91
11
93
97
94
47
«
19
91
24
39
91
13
22
74
1*
34
43. 4T
97.1
25. 4*
21.4
44.3
44.4
44.4
24. »
45.41
92.3.
91.4
97.1
14.9*
47.3
94.4.
47.97
47.4«
»9.4
94.4
14.4*
49.9
94.3*
43.4
94.4.
31.4.
3t.3T
24.4.
43.7
44.3
CCOC*APNIC OCMM UACMI TUOCS SO «0.
COOHOIHATCS ' CS STA
lAt LONC U4 «** VSCO
14.944'* 114.14* C 14 « 4.2 4
14.42
14.444
14.914
14.44*
14.419
49.149
17.412
14.944
34.432
42.134
14.422
1.443
34.444
14.4*4
14.431
9.14
•1.291
91.414
14.924
14.417
14.47*
44.4*1
34. *1*
34.441
51.1*4
7. 35
3*.4«4
34.*I4
47.9*4
N
«
«
N
S
N
S
H
U
N
S
N
N
N
N
S
13*.
11*.
114.
13*.
44.
4.
174.
11*.
*..
194.
177.
in.
11*.
11*.
11*.
194.
14 f
141 C
142 t
274 C
441 W
474 C
742 «
144 C
144 t
9*4 C
431 «
447 C
443 C
149 C
144 C
99 I
« 174.494 V
« 141.914 w
N 134.172 C
N
N
N
N
N
N
S
114
13*.
122
199 C
144 C
211 «
13*. 149 t
134.121 C
143.
12*.
233 «
.44 C
N 2*. 334 t
N
N
13*. 121 €
119.
.414 *
14
14
14
14
m
4
941
14
14
44
34*
922
14
• 14
14
144
11
91
14
14
14
19
14
14
1*
2*3
1*
1*
1*
C
C 4.9
C
C 4.4 4.4
0 4.4
C
4.4
C
C
0 4.7
4.4
• 9.4
C
C
e
t 1.9
« 4.9
« 4.4
C
f
C 4.2
C
C
G
C
•> 3.4
C
C
C
4.2
4.4
4.2
1.1
1.4
1.1
4.4
4.4
4.2
4.4
4.4
4/4
1.9
4.1
4.4
1.9
1.4
4.4
4.2
4.2
4.4
1.2
4.2
1.1
4.4
1.3
1.4
4.4
4.9
4
91
4
144
94
19
114
7
4
4
42
22
9
7
9
4
14
44
7
9
II
7
7
4
7
4
9
7
74
41 12 44 44.21 44.24* N 2*.4*4 14 C 4.4
41 12 14 43.27 4.14 S 134.44 114 7 4.1 1.2
41 12 5< 43.«X 44.372 « 3.441 It C 1.2
41 13 31 27.57 4.73 S 124.42 33 N 3.* 1.3
41 19 72-37.1- 21:424 N *4.*4« 12* • 4.7 1.1
41 14 43 4t.*7 42.4* S 45.73 14 C 4.4 4.7
41 14 17 34.4 34.414 N 13*.137 1* C 4.2
«(CICM. COMTHI4UTCO 4UCMITVOCS ANQ COUUCNTS
MCA* S. COASt OF MOKSHU. JAPAN. f«lt (II JUA) •* O«
•M (I JIM) •( A}tr«.
MCA« S. COASt Of HONSHU. JAPAN. r«l| (I jut.) «< A| 1
Ml (MM ««« m» O*kl 4M* 4M 0*k«4MS <•* 4W4T T«t4h4M.
NCA* S. COASf Of HONSHU. JAPAN, foil
(II JMA) 4t H4«4U«<« 4M 1«ky«: (I 4MA) «t
K •«•««<•< I -k«. *
CMIU-40KVIA •MOC4 «CCIOM
riUHCt. «l 9.9 («CM). 9.1 (CCC).
flJI I91AMO9 HCCIOM
MCA« s. COASt or MOHSMg. JAPAN. r«it
-------
J2(r 160* 160* 120*
«>•
40* 80*
80*
Figure 2.1 The Six Major Tectonic Plates and Their Approximate Linear Velocity
Vectors (adapted from Park, 1983).
19
-------
SAN FRANCISCO
1906
Figure 2.3 Isoseismal Contours for Intra-Plate vs. Edge-Plate Events of Similiar
Magnitude (Nuttli, 1981)
21
-------
4O F
30
-88 -86 -84 -82 -80 -78 -76
Longitude (Deg.)
Figure 2.4 Epicenters for Earthquakes M £ 2.5 in the Southeastern United States (July
1977 - December 1984) (Sibol et al., 1984).
22
-------
HMN ZONE Of
r*uu Evidence
K
TOTAL VIOTV OFTAUIT ZONC 78 FtCT
Figure 2.5 Characteristics of Hayward Fault as Exposed in Five Trenches at Fremont
Site (duff etal., 1972).
23
-------
o
V
«!
*•
™
1-
II
o
II
<
•
<
IU
4
i
i
3
1
1
't
^
<
i
4
u.
4
1
1
_J
1
z
u
u
R
j
CO
J
<12
o o
z <
2?
IU
3
i i
xi IU
IU Z
tt. O
»»l
o IU O
o —
iu r> z
o o
< O 1-
Cf UJ
t- Z UJ
•™* u.
1- <
-J 3C ^^
U. .•-,-_'__-.
l_
D; y
IU /
s /
J
J
/
/
/
/
/
/
/ A
/ "
/ "
/ *
/ ^
/''^
' * ^
o {P-
/
-------
SECTION 3
258.14 SEISMIC IMPACT ZONES:
USGS PROBABILISTIC BEDROCK ACCELERATION
Subtitle D provides for two alternative methods to determine the maximum (peak) horizontal
acceleration (MHA) for design of MSWLF's. The prescriptive method of determining the MHA
is from a seismic hazard map depicting the peak horizontal ground acceleration in lithified earth
(bedrock) with a probability of 90 percent (or greater) of not being exceeded in a 250 year period.
Either USGS Map Sheet MF-2120 (USGS, 1982; USGS, 1990), as presented in Figure 1.1 and
used to define the extent of seismic impact zones in the United States, or an equivalent map
acceptable to the director of a USEPA-approved state or tribal regulatory program may be used.
Alternatively, the MHA may be based upon a site specific seismic hazard analysis. The details
of what constitutes an acceptable site specific hazard analysis are not provided in Subtitle D.
Rather, these details are left to the discretion of the director of a USEPA-approved state or tribal
regulatory program. Many states simply provide for site specific analysis that determine the peak
horizontal ground acceleration in lithified earth with a 90 percent (or greater) probability of
exceedance in 250 years. Figure 3.1 outlines the steps required for such an analysis. Other states
allow for use of deterministic analysis to determine the largest or most damaging earthquake
expected to impact the site.
Many experts consider the use of site specific analyses preferable to use of generic seismic hazard
maps for assessing the peak ground acceleration for engineering analyses due to the ability to
achieve a greater degree of precision and to incorporate up-to-date information on regional
seismology and tectonics in a site specific analysis (Anderson and Kavazanjian, 1995). Other
experts maintain not only the superiority of site specific analyses over seismic hazard but also the
superiority of deterministic analyses over probabilistic seismic hazard evaluation (Krinitzsky,
1993). Discussion of these issues is beyond the scope of this guidance document.
The USGS probabilistic seismic hazard map presented in Figure 1.1 provides the prescriptive
means of determining the MHA for MSWLF. The latest available version of the map should be
used in the analyses. If knowledge and understanding of local and regional seismology and
geology or attenuation of earthquake ground motions have changed sufficiently since development
of the map to invalidate the map, or if it is believed that the map is otherwise inappropriate, a site-
specific analysis may be warranted. The details of an acceptable site specific analysis are left up
to the discretion of a USEPA-approved state or tribal regulatory program. However, considering
the relative low probability and large return period associated with the peak acceleration evaluated
from the USGS map, in most situations landfills designed using the peak acceleration from the
25
-------
most recent version of the map will afford a high degree of protection to the environment against
earthquakes reasonably expected over the life and post-closure care period of the MSWLF.
The USGS map presents the estimated peak ground acceleration for a hypothetical bedrock
outcrop at a project site. If bedrock is not present at or near the ground surface, the peak
acceleration from the USGS map may need to be modified to account for local site conditions. The
primary difficulty associated with the USGS or other seismic hazard map is that such maps
typically do not provide information on the magnitude or duration of the earthquake associated
with the map acceleration values. In fact, the acceleration values provided on such maps are
typically composed of contributions of earthquakes of many different magnitudes at many different
distances. For most geotechnical analyses, peak acceleration and magnitude are necessary.
Distance and/or duration may also be required for certain geotechnical evaluations.
This section of the guidance document provides background on the methodology used to generate
the USGS seismic probability map and discusses interpretation of the acceleration value obtained
from the map in order to obtain site-specific seismicity parameters for design.
3.1 Development of Design Earthquake
The USGS map is shown on Figure 1.1 in a reduced size format. The original map generated
by USGS is sufficiently large that individual counties within the states are shown for ease in
locating a particular landfill site. Selection of a peak horizontal ground acceleration from this map
is a straight forward process. However, association of a magnitude with this peak acceleration
requires interpretation and judgement.
An acceleration value from the USGS maps for any particular site is composed of contributions
from a family of earthquakes of different magnitudes and distances. Figure 3.2 (Moriwaki et al.,
1994) shows the distribution of magnitudes and distances from a hypothetical probabilistic seismic
hazard analysis for a 10 percent probability of exceedance (90 percent probability of not being
exceeded) over a 50 year exposure period (this corresponds to a 475 year return period).
Selection of a representative magnitude from this data might be based upon either a 90 percentile
criterion or the mean (expected) value at the discretion of the design engineer and regulatory
agency.
The information on the distribution of magnitudes is generally not available for USGS or other
regional seismic hazard studies, and most common seismic hazard programs must be modified to
yield this data. As an alternative, the maximum magnitude assigned to the seismic source zones
which contribute to the seismic hazard (the zone the site is in plus all adjacent zones) may be
conservatively taken as the representative magnitude. The source zones used to develop the 1982
26
-------
USGS map are presented in Figure 3.2. No supplemental documentation was presented with the
1990 map. Maximum magnitudes are presented in Table 3.1 for the USGS source zones shown
in Figure 3.3, and Figure 3.4 presents maximum magnitudes for much of the central United States
from a more recent study (Johnston and Nava, 1994). Knowledge and understanding of seismic
source zones is continually advancing. The prudent investigator should consult current sources
of information on local and regional seismicity to identify the source zones impacting the site and
the maximum magnitudes assigned to these source zones.
In performing a seismic performance analysis for a MSW landfill, it should be recognized by the
engineer and regulator that the peak ground acceleration at the site is not always the acceleration
associated with the most damaging earthquake. Large magnitude earthquakes at large distances
can generate ground motions at a site that are of lower intensity but greater damage potential than
a small magnitude nearby event associated with the MHA. Use of the maximum magnitude from
all contributing source zones as the magnitude associated with the MHA combined with the low
probability occurrence and large return period associated with the MHA will generally provide
a design event of sufficient damage potential to provide a high degree of environmental protection
over its' active life and post closure care period.
It is anticipated that use of the maximum magnitude event from all contributing source zones in
combination with the USGS map MHA will often produce a very conservative assessment of the
design event. However, in some situations, most notably when the site is on the fringe of a major
seismic source zone capable of generating a great earthquake (e.g., San Andreas, New Madrid,
Charleston, Cascadia Subduction Zone), the large distant event with a lower PGA may be the
most damaging earthquake and an event-specific analyses for the MHA associated with the great
earthquake may be warranted in addition to the use of the MHA from the seismic hazard map.
If such an event-specific analysis is conducted, the magnitude from the great earthquake source
zone need not be considered in evaluating the magnitude associated with the USGS map
acceleration.
3.2 Interpretation of Peak Bedrock Accelerations
The attenuation relationships used to establish the USGS seismic probability maps are based on
ground motions recorded at bedrock sites. Bedrock is commonly defined in engineering practice
as material having a shear wave velocity greater than 2,500 feet per second (750 meters per
second). This is referred to as lithified earth within Subtitle D. Lithified earth is defined in
Subtitle D as all rock, including all naturally occurring and naturally formed aggregates or masses
of minerals or small particles of older rock that formed by induration of loose sediments.
Lithified earth does not include man-made materials such as fill, concrete and asphalt, or
unconsolidated earth materials, soil, or regolith (saprolites) lying at or near the ground surface.
27
-------
It is important to realize that the accelerations presented on the USGS maps are not the peak
ground surface accelerations unless bedrock is exposed at the ground surface. Section 4.1 of this
guidance document reviews methods for calculating the peak ground surface acceleration based
on the site specific subgrade profile that exists above the top of lithified earth (rock) and the peak
bedrock acceleration from the USGS map.
The peak acceleration is only one characteristic of the earthquake ground motion at a site. The
damage potential of seismically-induced ground motions also depends upon the duration of the
motion, the frequency content of the motion, and the intensity of the motion at times other than
when the peak acceleration occurs. Acceleration, velocity, and displacement time-histories
recorded at the top of the Oil landfill in Los Angeles during the 17 January 1994 Northridge
earthquake (Hushmand Associates, 1994) are shown on Figure 3.5. Note that the peak
acceleration occurs only once during the record and that motions approaching the peak
acceleration exist for only a small fraction of a second. Use of this peak acceleration for
traditional geotechnical stability analyses is very conservative in most cases. Section 6.2 of this
guidance document discusses the reduction of this peak ground surface acceleration to an
equivalent pseudo-static acceleration for use in slope stability analyses.
3.3 References
Algermissen, S.T., Perkins, D.M. Thenhaus, P.C., Hanson, S.L., and Bender, B.L. (1982),
"Probabilistic Estimates of Maximum Acceleration and Velocity in Rock in the Contiguous United
States," U.S. Geological Survey Open-File Report 82-1033, 99 p.
Anderson, D.G., and Kavazanjian, E., Jr. (1995), "Performance of Landfills Under Seismic
Loading," Proc. Third International Conference on Recent Advances in Geotechnical Earthquake
Engineering and Soil Dynamics, University of Missouri, Rolla, Vol. 3, 2-7 April.
Bonilla, M.G., Mark R.K., and Lienkaemper, J.J. (1984), "Statistical Relations Among
Earthquake Magnitude, Surface Rupture Length and Surface Fault Displacement, Journal of
Geophysical Research, Vol. 74, No. 6. pp. 2379-2411.
Boore, D.M., and Joyner, W.B. (1994). "Prediction of Ground Motion in North America," Proc.
of the Seminar on New Developments in Earthquake Ground Motion Estimation and Implication
for Engineering Design Practice, Applied Technology Council Publication No. ATC 35-1,
Redwood City, California, pp. 6-1 - 6-41.
Cornell, C.A. (1968), "Engineering Seismic Risk Analysis," Seismological Society of America
Bulletin, V. 58, pp. 1583-1606.
28
-------
Hanks, T.C., and Kanamori, H. (1979), "A Moment Magnitude Scale," Journal of Geophysical
Research, Vol. 84, No. B5, pp. 2348-2350.
Heaton, T.J., Tajima, F., and Mori, A.W. (1986), "Estimating Ground Motions Using Recorded
Accelerograms" Surveys in Geophysics, Vol. 8, pp. 25-83.
Hushmand Associates (1994), "Landfill Response to Seismic Events," Report prepared for the
USEPA Region IX, Hushmand Associates, Laguna Niguel, California.
Idriss, I.M. (1985), "Evaluating Seismic Risk in Engineering Practice," Proc. Eleventh
International Conference on Soil Mechanics and Foundation Engineering, San Francisco,
California, Vol. 1, pp. 255-320.
Johnston, A.C., and Nava, S.J. (1994), "Seismic Hazard Assessment in the Central United
States," Proc. Seminar on New Developments in Earthquake Ground Motion Estimation and
Implications for Engineering Design Practice, Applied Technology Council, ATC35-1, Redwood
City, California, pp. 2-1 - 2-12.
Krinitzsky, E.L. (1993), "Earthquake Probability in Engineering - Part 2: Earthquake Recurrence
and Limitations of Gutenberg-Richter b-values for the Engineering of Critical Structures,"
Engineering Geology, Vol. 36, pp. 1-52.
Moriwaki, Y., Tan, P. and Somerville, P. (1994); "Some Recent Site-Specific Ground Motion
Evaluations - Southern California Examples and Selected Issues," Proc. Seminar on New
Developments in Earthquake Ground Motion Estimation and Implications for Engineering Design
Practice, Applied Technology Council ATC35-1, Redwood City, California, pp. 14-1 - 14-25.
Nuttli, O.W. (1981), "Similarities and Differences Between Western and Eastern United States
Earthquakes, and their Consequences for Earthquake Engineering," Earthquakes and Earthquake
Engineering: The Eastern United States, Vol. 1, Assessing the Hazard - Evaluating the Risk, I.E.
Beavers, Ed., Ann Arbor Science Publishers, Inc., Ann Arbor, Michigan, pp. 25-51.
USGS (1982), "Probabilistic Estimates of Maximum Acceleration and Velocity in Rock in the
Contiguous United States, United State Geological Survey, Open-File Report 82-1033.
Wesnousky, S.G. (1986), "Earthquakes, Quaternary Faults and Seismic Hazard in California,"
Journal of Geophysical Research, Vol. 91, No. B12, pp. 12587-12631.
29
-------
Woodward-Clyde Consultants (1979), "Evaluation of Maximum Earthquake and Site Ground
Motion Parameters Associated with the Offshore Zone of Faulting, San Onofre Nuclear
Generating Station," Report prepared for Southern California Edison Company, 241 p.
Wyss, M. (1979), "Estimating Maximum Expectable Magnitude of Earthquakes from Fault
Dimensions," Geology Vol. 7, pp. 336-340.
30
-------
Table 3.1: Parameters for Seismic Source Zones (USGS, 1982).
Zone
No.*
pOOl
p002
p003
pOOA
pOOS
p006
p008
p009
pOlO
pOll
p012
p013
pO!4
p015
p016
pO!7
p018
p019
cOOl
c002
c003
0004
c005
c006
0007
0008
c009
cOlO
cOll
c012
c013
c014
c015
c016
cO!7
0018
c019
c020
c021
c022
c023
0024,
c025
o026
No. of Modified
Mercalll Maxioua
Intensity V'«
per year
0.11010
0.43510
0.12440
0.34840
0.12390
0.02831
0.01642
0.20850
0.45200
0.96370
0.37090
0.69020
0.10940
0.34480
0.04926
0.87860
0.18810
0.04090
0.62770
0.15700
0.31960
0.31960
0.04843
0.15700
0.15700
0.04740
0.04843
0.18190
0.77010
0.19050
0.35840
0.91990
1.49200
0.22560
0.02760
1.09200
0.31980
0.19280
0.10880
0.02422
0.11650
1.97000
0.05085
0.09145
b
-0.40
-0.40
-0.54
-0.62
-0.62
-0.62
-0.42
-0.28
-0.28
-0.28
-0.28
-0.28
-0.42
-0.42
-0.42
-0.28
-0.54
-0.54
-0.42
-0.4"2
-0.42
-0.42
-0.42
-0.42
-0.42
-0.42
-0.42
-0.42
-0.42
-0.42
-0.42
-0.66
-0.45
-0.51
-0.48
-0.49
-0.42
-0.42
-0.42
-0.42
-0.37
-0.43
-0.55
-0.55
Maximum
Magnitude
M**
7.3
7.3
7.3
7.3
7.3
7.3
7.3
7.9
7.9
7.9
7.9
7.9
7.3
7.3
7.3
7.9
7.3
7.3
7.3
7.3
7.3
7.3
6.1
7.3
7.3
6.1
6.1
6.1
7.3
7.3
7.3
7.9
7.9
7.9
7.3
7.3
6.7
6.1
6.1
6.1
7.9
8.5
7.3
7.3
31
-------
Table 3.1: (continued)
Zone
No.*
c027
c028
c029
efl30
c031
C032
c033
C034
c035
d336
c037
d)38
c039
CD40
cQ41
001
002
003
004
005
006
007
008
009
010
Oil
012
013
014
015
016
017
018
019
020
022
023
024
025
026
027
029
030
031
No. of Modified
Hercalli Maximum
Intensity V's
per year
0.03437
0.13010
0.02350
0.03630
0.47580
0.55190
0.23070
0.67120
0.02325
0.35220
0.81950
0.82680
0.35810
0.15820
0.08448
0.22700
0.03600
0.08800
0.22700
0.09100
0.13500
0.41900
0.21100
0.19400
0.20800
0.55100
0.34900
0.05500
0.49000
0.01800
0.14600
0,69300
0.26100
0.11717
1.84900
0.19600
0.15350
0.27400
0.16800
0.47700
0.1 1100
1.31900
0.58800
1.82685
b
-0.37
-0.37
-0.37
-0.42
-0.51
-0.45
-0.37
-0.51
-0.60
-0.59
-0.51
-0.54
-0.45
-0.42
-0.37
-0.73
-0.73
-0.73
-0.54
-0.73
-0.73
-0.73
-0.73
-0.54
-0.54
-0.64
-0.64
-0.64
-0.73
-0.73
-0.73
-0.59
-0.54
-0.54
-0.64
-0.64
-0.54
-0.64
-0.64
-0.64
-0.64
-0.64
-0.64
-0.54
Maxima
Magnitude
M**
7.3
7.3
7.3
6.7
6.7
7.9
7.9
7.9
7.3
6.7
6.1
7.9
7.9
6.1
7.9
7.3
7.3
6.1
7.3
7.3
7.3
7.3
6.1
6.1
7.3
7.3
7.3
7.3
7.3
6.7
6.1
7.3
7.3
7.3
7.3
6.1
7.3
7.3
6.1
6.1
5.5
7.3
7.3
7.3
32
-------
Table 3.1: (continued)
Zone
No.*
032
033
034
035
036
037
038
039
040
041
042
043
044
045
046
047
048
049
050
051
052
053
054
055
056
057
058
059
060
061
062
063
064
065
066
067
068
069
070
071
072
073
074
075
No. of Modified
Mercalli Maxima
Intensity V's
per year
0.48114
0.08557
0.6 2380
0.20070
0.01800
0.05100
0.80600
0.12000
0.29100
0.24400
0.01800
0.04600
0.11300
0.45600
0.01274
0.00427
0.00329
0.01663
0.17000
0.01706
0.19000
0.03600
0.01800
0.67300
0.17700
0.66200
0.19800
0.19200
0.03600
0.08900
0.03600
0.12900
0.34400
0.15200
0.01800
0.07715
0.02894
0.00588
0.03552
0.01176
0.02026
0.02353
0.00270
0.06510
b
-0.54
-0.54
-0.54
-0.54
-0.58
-0.58
-0.58
-0.58
-0.58
-0.73
-0.73
-0.73
-0.73
-0.73
-0.73
-0.73
-0.73
-0.73
-0.73
-0.73
-0.58
-0.58
-0.58
-0.58
-0.58
-0.58
-0.58
-0.58
-0.58
-0.58
-0.58
-0.58
-0.58
-0.58
-0.73
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
Maximum
Magnitude
M**
6.1
6.1
7.3
7.3
6.1
7.3
7.3
7.3
7.3
7.3
6.1
7.3
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6.1
7.3
7.3
6.1
7.3
6.1
7.3
7.3
6.1
6.1
7.3
6.1
6.1
7.3
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6,1
6.1
33
-------
Table 3.1: (continued)
Zone
No.*
076
077
078
079
080
081
082
083
084
085
086
087
088
089
090
091
092
093
094
095
096
097
098
099
100
101
102
103
104
106
107
108
109
110
111
112
113
114
115
116
117
118
No. of Modified
Mercalli Maxima
Intensity V's
per year
0.14742
0.03469
0.04389
0.03082
0.02987
0.02044
0.03552
0.00996
0.04117
0.03802
0.04626
0.29865
0.09703
0.15689
0.06103
0.00644
0.02661
0.02680
0.10835
0.05901
0.02675
0.01156
0.01215
0.24830
0.42290
0.18720
0.09532
0.33150
0.05544
0.01952
0.19100
0.29390
0.10650
0.30220
0.32430
0.01532
0.07432
0.00754
0,05834
0.06783
0.03950
0.01334
b
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.46
-0.50
-0.50
-0.50
-0.50
-0.5Q
-0.50
-0.50
-0.50
-0.50
-0.50
-0.50
-0.50
-0.50
-0.50
-0.50
-0.50
-0.50
-0.50
-0.50
Maxima
Magnitude
M**
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6.1
8.5
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6.1
6.1
7.3
7.3
7.3
7.3
7.3
7.3
6.7
7.3
6.7
7.9
7.9
7.9
6.7
6.7
6.7
7.3
6.7
7.3
7.3
*The zones are shown in Figure 3.2
**See text for definition of M
34
-------
(A)
Source 1 (FAULT)
Source 2
(POINT)
o»
o
(B)
MAGNITUDE
Source 3 (MIXED)
C
O
o
v.
JOJ
o
o
8.5
Distance
A!
VI
<
Al
0.2
Acceleration
Acceleration
Figure 3.1 Basic Elements of the USGS Probabilistic Hazard Calculations: (a) Typical
Source Areas and Grid of Points at Which the Hazard is to be Computed;
(b) Statistical Analysis of Seismicity Data and Typical Attenuation Curves;
(c) Cumulative Conditional Probability Distribution of Acceleration; (d)
The Extreme Probability, F^., (a) for Various Accelerations and Exposure
Times (T) (USGS, 1982).
35
-------
475-Year Average Return Period
c
JO
53
.a
o
o
5.0 5.5 6.0 6.5 7.0 7.5 8.0
Magnitude
0-15 km
Figure 3.2 Contribution of Various Magnitudes and Distances to the Seismic Hazard
(Moriwaki et al., 1994).
36
-------
u>
-J
Figure 3.3 Seismic Source Zones in the Contiguous United States (USGS, 1982).
-------
so*
-45*
Superior Oat on
A M»X5.O
Background
B Mb5.0-5i5
jjST
MK 6.0
Reelfoof
*•• f 10
Okl. Aul.
T
Mb6.8
to
Mh
-------
o
I
LU
UJ
6
U
2
0
£ -2
tn
»—< —y
o
-6
0
-30
10
10
10
20
30
20
30
110
20 30 UO
TIME - seconds
50
60
50
60
50
60
Figure 3.5 Time-Dependant Fluctuations in Seismic Ground Response Parameters (17
January 1994 Northridge, California Earthquake, OH Site, Longitudinal
Component) (Hushmand Associates, 1994).
39
-------
SECTION 4
258.14 SEISMIC IMPACT ZONES:
SITE SPECIFIC SEISMIC DESIGN GROUND MOTION
The USGS map discussed in the previous section provides values for the peak ground acceleration
of a hypothesized bedrock outcrop at a MSW landfill site. This section of the guidance document
discusses methods for calculation of: (1) a peak acceleration in the free field at the ground surface
at the project site that reflects the soil stratigraphy and (2) a peak acceleration at the top of the
landfill that reflects the properties of the waste. These accelerations are used in later sections of
this guidance document in evaluation of the seismic response of the landfill waste mass, the
seismic performance of the liner and cover systems, and subgrade liquefaction potential.
Qualitative reports of the influence of local soil conditions on the intensity of shaking and on the
damage induced by earthquake ground motions date back to at least the 1906 San Francisco
earthquake (Wood, 1908). Reports of localization of areas of major damage within the same city
and of preferential damage to buildings of a certain height within the same local area from the
Mexico City earthquake of 1957, the Skopje, Macedonia earthquake of 1963, and the Caracas,
Venezuela earthquake of 1967 focused the attention of the engineering community on local soil
effects.
Back-analysis by Seed (1975) of accelerograms from the magnitude M 5.7 San Francisco
earthquake of 22 March 1957, presented in Figure 4.1, demonstrate the influence of local soil
conditions on site response. Peak accelerations and the frequency contents of ground motions
measured at six sites approximately the same distance from the earthquake source were dependent
on the soil profile beneath each specific site.
Figure 4.1 shows peak acceleration, the acceleration and velocity response spectra, and soil
stratigraphy data at the six San Francisco sites from the 1957 earthquake. A response spectrum
presents the maximum response of a damped single degree-of-freedom (SDOF) linear elastic
system to the accelerogram recorded at a site. The maximum response of the SDOF system is
calculated for a range of system natural frequencies to plot the response spectrum. Response
spectra are typically calculated for several levels of system damping, as shown on Figure 4.2.
Acceleration data generated in response spectra analysis is commonly plotted on the tripartite plot
shown on Figure 4.3. In addition to peak acceleration, the tripartite presentation also provides
approximate values of peak velocity and peak displacement for the response of the SDOF.
At the sites shown hi Figure 4.1, the local soil deposits attenuated the peak ground acceleration
by a factor of approximately two compared to the bedrock sites. However, the acceleration
40
-------
response spectra clearly show amplification of spectral accelerations at longer periods (periods
greater than 0.25 sec). If the bedrock motions had greater energy at these longer periods, a
characteristic of larger magnitude events and of events from a more distant source, or if the
natural period of the local soil deposits more closely matched the predominant period of the
bedrock motions, amplification of the peak acceleration could have occurred at the soil sites.
Amplification of long period bedrock motions by local soil deposits is now accepted as an
important phenomenon that can exert a significant influence on the damage potential of earthquake
ground motions. Significant structural damage has been attributed to amplification of both peak
acceleration and spectral acceleration by local soil conditions. Amplification of peak acceleration
occurs when the resonant frequency of the soil deposit is close to the predominant frequencies of
the bedrock earthquake motions (the frequencies associated with the peaks of the acceleration
response spectra). The resonant frequencies, /„, of a soil layer (deposit) of thickness H can be
estimated as a function of the average shear wave velocity of the layer, Vs, using the following
equation:
v. (2» *) .. ,
where/; is the resonant frequency for the first mode of vibration,/2 is the resonant frequency for
the second mode of vibration, f3 is the resonant frequency for the third mode of vibration, and so
on. At most soil sites, amplification of seismic motions is most important for the first
(predominant) mode of vibration and rapidly decreases in significance with increasing mode
number.
Spectral amplification may occur at soil sites in any earthquake at frequencies around the resonant
frequency of the soil deposit. Spectral amplification causes damage when the resonant frequency
of the soil deposit matches the resonant frequency of the structure. Some of the most significant
damage in recent earthquakes (e.g., building damage in Mexico City in the 1985 earthquake and
damage to freeway structures in the Loma Prieta earthquake of 1989) has occurred in situations
where the predominant frequencies of the bedrock motions and the resonant frequencies of both
the local soil deposit and the overlying structure all fell within the same range.
Observations of ground motions generated in recent earthquakes at the Oil landfill, a solid waste
landfill in Los Angeles composed of both industrial and municipal wastes, have demonstrated that
amplification of both spectral acceleration and peak acceleration can occur at the top of solid waste
landfills. Anderson et al. (1992) report spectral amplification of greater than 10 at Oil for low
amplitude (less than 0.1 g) ground motions from small magnitude (less than M 5.0) earthquakes.
41
-------
Hushmand Associates (1994) report a peak horizontal acceleration amplification factor of 3.0 at
Oil during the M 7.4 Landers earthquake in 1992.
Considering the landfill facility as an engineered structure built upon a local soil deposit, there
are clearly two different sources of local site effects that must be considered in a seismic impact
analysis. First, the influence of the local soil conditions on the bedrock motions must be evaluated
to determine the free field ground surface motions at the project site. Second, the influence of the
landfill on the free field ground surface motions must be evaluated. While it is convenient
conceptually to separate these two effects, in practice they may be inter-dependent and a coupled
analysis of the interaction between the response of the foundation soil and the response of the
landfill may be warranted.
This section of the guidance document presents simplified and detailed methods for evaluating
both the free field ground response and the response of the landfill mass. The free field ground
motions are used to evaluate the liquefaction potential of the foundation. The response analysis
of the landfill mass provides input for seismic performance analyses of the landfill liner and cover
systems.
4.1 General Methodology
The influence of local soil conditions on seismic ground motions is usually addressed using one-
dimensional site response analyses. Conventional one-dimensional site response analyses are
based upon the assumption of a horizontal shear wave propagating vertically upwards through
horizontal soil layers of infinite lateral extent. The influence of vertical motions, compression
waves, and laterally non-uniform soil conditions are typically not accounted for in a one-
dimensional site response analysis. Similarly, geotechnical engineering analyses of liquefaction
potential and seismic stability consider only the horizontal component of the seismic motions.
This reliance solely on the horizontal component is consistent with common design and code
practices.
The most common analytical method used for one-dimensional site response analyses is the
equivalent linear method, wherein a layered vertical soil column is treated as a linear visco-elastic
material characterized by an elastic modulus and a viscous damping ratio. To account for the non-
linear, strain-dependent behavior of soil, the equivalent linear modulus and damping ratio are
evaluated from the modulus and damping measured in uniform cyclic loading at the
"representative" shear strain. Based on comparison of observed seismic site response with site
response predicted using equivalent linear analysis, the representative shear strain is usually taken
as 65 percent of the maximum shear strain calculated in the site response analysis. Because the
maximum shear strain is not known prior to the start of an analysis, equivalent linear response
42
-------
analyses are performed in an iterative manner. The maximum shear strain from one run is used
to evaluate the equivalent modulus and damping for the next run and continuing to convergence.
Input to one-dimensional equivalent linear site response analyses typically includes the shear wave
velocity and mass density or small-strain shear modulus for each soil layer, curves relating the
representative shear strain to a modulus reduction factor and the fraction of critical damping for
each soil type (modulus reduction and damping curves), the representative shear strain factor (the
fraction of the maximum shear strain assumed to correspond to the representative shear strain) and
an input acceleration-time history. Other input parameters include the density and shear wave
velocity of the underlying bedrock. The acceleration-time history may be input as the motion at
a hypothetical bedrock outcrop or at the bedrock-soil interface at the base of the soil column.
Results of the analysis provide shear stress- and acceleration-time histories for each layer within
the soil profile.
An alternative to the equivalent linear method of site response analysis is truly non-linear site
response analysis (Lee and Finn, 1978; Matasovi and Vucetic, 1993). In a truly non-linear
analysis, the actual hysteretic stress-strain behavior of each element of soil (or waste) is calculated
in the time domain. Equivalent linear analysis are typically performed in the frequency domain,
employing the principal of superposition to calculate the time history of ground motions. Non-
linear site response analyses require a description of the hysteretic stress strain behavior of the soil
(or waste), the mass density profile of the material, and an input acceleration time history. Truly
non-linear site response analyses hold the promise of a more accurate representation of the seismic
behavior of soil deposits and solid waste landfills. However, at the present time, truly non-linear
site response analyses are still primarily a research tool and have yet to be widely employed in
engineering practice.
4.1.1 Simplified Analysis
Whereas structural analyses typically require information on the spectral content of ground
motions, and thus require a complete time history to characterize the design motion, geotechnical
analyses frequently only require knowledge of either the peak ground acceleration or the peak
ground acceleration and the earthquake magnitude. Several investigators have related the peak
ground acceleration from a hypothetical bedrock outcrop, such as presented on the USGS maps,
to the peak ground acceleration at a specific site as a function of the local soil conditions based
upon the results of one-dimensional site response analysis and observations of ground motions
during earthquakes. The top plot on Figure 4.4 shows an early relationship developed by Seed
and Idriss (1982) for a variety of local soil conditions. This plot was developed using SHAKE,
a computer program for equivalent linear one-dimensional site response analyses developed at the
University of California, Berkeley (Schnabel et al., 1972).
43
-------
Experience from recent earthquakes has shown that the curves in the top plot of Figure 4.4 can
significantly under-predict site amplification effects in many situations. The plot on the bottom
of Figure 4.4 shows a recent curve developed by Idriss (1990) for soft soil sites. This plot was
developed from both SHAKE analysis and from field observations of soft soil site response in two
recent earthquakes.
Observations of the response of the Oil landfill, Los Angeles, in recent earthquakes (Hushmand
Associates, 1994) and the results of truly non-linear one-dimensional seismic response analyses
of landfills (Kavazanjian and Matasovi, 1995) indicate that the Idriss (1990) soft soil-site
amplification curve may also provide an appropriate representation of the potential for typical peak
acceleration amplification at solid waste landfills. Data obtained at the Oil landfill during four
recent earthquakes is plotted in Figure 4.5 along with the soft soil site field data and recommended
curve from Idriss (1990). Also plotted on this figure are the results of non-linear analyses of
landfill seismic response performed by Kavazanjian and Matasovi (1995) using waste parameters
backfigured from strong motion records obtained at the Oil landfill in the 17 January 1994 M 6.7
Northridge earthquake (peak acceleration at the landfill crest equal to 0.24 g).
Some of the non-linear landfill response analyses results plotted on Figure 4.5 at 0.3 g and 0.5 g
bedrock acceleration fall significantly above the Idriss (1990) curve. However, the results that
fall above the Idriss curve are from low amplitude (less than 0.1 g) accelerograms recorded at
large distances from the earthquake source (greater than 50 kilometers) that were scaled up to
large accelerations representative of near field conditions. Therefore, the large amplification
factors computed for these cases may not be representative of the amplification potential from real
earthquakes. On this basis, Kavazanjian and Matasovi (1995) concluded that the Idriss (1990) soft
soil amplification curve provides a reasonable representation of the average peak acceleration
amplification potential at the top of solid waste landfills.
Figure 4.6, from Singh and Sun (1995), present the results of their theoretical analyses for the
amplification potential of a 100 ft (30 m)- high refuse fill along with a summary of the upper
bound for observations of amplification at the crest of earth dams in earthquakes by Harder
(1991). These investigators suggest that the observational data on earth dams may also provide
an upper bound on the amplification potential of waste fills. The earth dam curve corresponds
closely to the upper range of the analytical results of Kavazanjian and Matasovi (1995) presented
in Figure 4.5.
The soft soil site curve developed by Idriss and presented in Figure 4.5, the analytical data
developed by Kavazanjian and Matasovi (1995) and presented in Figure 4.5, and the observatinal
data presented in Figure 4.6 may be used in a three-step simplified procedure developed by
GeoSyntec (1994) to perform a simplified site response analyses for the purpose of adjusting the
44
-------
peak acceleration from the USGS map (or from a site-specific analysis for the MHA in lithified
earth) for the influence of local soil conditions (to obtain the free field peak acceleration at a
project site) and for the influence of the landfill (to obtain the peak acceleration at the crest of the
landfill). The three-step procedure is as follows:
Step 1: Classify the Site. Classify the site as special study, soft, medium stiff, stiff, or rock on
the basis of the average shear wave velocity for the top 30 meters (100 feet) of soil and
the following table (Borcherdt, 1994):
CLASSIFICATION AVERAGE SHEAR WAVE VELOCITY
Special Study Less than 100 m/s (330ft/s)
Soft 100 to 200 m/s (330 to 660 ft/s)
Medium Stiff 200 to 375 m/s (600 to 1,230 ft/s)
Stiff 375 to 700 m/s (1,230 to 2,300 ft/s)
Rock Greater than 700 m/s (2,300 ft/s)
Note that special study soils also include liquefiable soils, quick and highly sensitive
clays, peats, highly organic clays, very high plasticity clays (PI>75%), and soft soil
deposits more than 37 meters (120 feet) thick.
Step 2: Estimate the Free field Acceleration. Estimate the potential amplification of the
bedrock motions by the local soil deposit based upon the soil profile classification. For
soft soils, use the curve in Figure 4.5 recommended by Idriss (1990) to estimate the
free field peak ground acceleration from the peak bedrock acceleration. For medium
stiff soils, use an acceleration equal to the average of the rock site acceleration and the
soft soil site acceleration from Idriss' curve in Figure 4.5 for peak bedrock
accelerations less than or equal to 0.4 g. For medium stiff soils when the peak
bedrock acceleration exceeds 0.4 g and for stiff sites for all acceleration levels, assume
the free field peak ground acceleration at the site is equal to the peak bedrock
acceleration. For Special Study soil sites, Figure 4.5 should not be used. Instead, site
specific seismic response analyses such as those described in the next section of this
guidance document should be conducted.
Step 3: Estimate the Peak Acceleration at the Top of the Landfill. Estimate the potential
amplification of the peak acceleration of the landfill mass using the analytical data in
Figure 4.5 and the earth dam amplification curve in Figure 4.6. The decision as to
whether to use the upper bound of the analytical data and/or the earth dams
observations or to use a value closer to the median of the analytical data in Figure 4.5
45
-------
is a matter of engineering judgment. As a general rule, if the predominant period of
either the design earthquake or the foundation soil matches the predominant period of
the landfill, the upper bound data should be used. The predominant period of the
landfill and the foundation may be evaluated using equation 4.1. For relatively thin
fills (less than 33 ft (10 m) founded on soft soils or subject to long period motions
from distant earthquakes and for thick fills (over 165 ft (50 m)) founded on rock
subject to high frequency motions from nearby earthquakes, the median analytical data
(corresponding roughly to the Idriss soft soil curve) may be appropriate. The free field
ground acceleration developed in Step 2 is used in place of the peak bedrock
acceleration on the abscissa of Figures 4.5 and 4.6, and the acceleration at the top of
the landfill is obtained from the ordinate of the appropriate figure.
The three-step procedure presented above is a simplified, decoupled analysis that ignores
interaction between the waste mass and the ground. Analyses of the coupled response of landfills
and foundation soils indicates that this simplified, decoupled analysis will yield a conservative
upper bound estimate of the combined amplification potential of a landfill and its foundation
(Bray, etal., 1995; GeoSyntec, 1994).
The peak acceleration at the top of the landfill estimated in Step 3 may be used in seismic
performance analyses of the landfill cover and surface water drainage systems and in evaluation
of other facilities constructed on top of the landfill (e.g., flare station or storage tanks). The
acceleration calculated in Step 3 is not, however, the appropriate peak acceleration for use in
seismic stability and deformation potential calculations of the waste mass. For seismic stability
and deformation potential evaluations of the waste mass, the average acceleration of the assumed
failure mass, and not the acceleration at the top of the landfill, is the relevant response quantity,
as the average acceleration is directly proportional to the seismically-induced inertia forces and
to the seismic shear stresses induced at the base of the failure mass (Repetto et al., 1993).
Makdisi and Seed (1978) developed a "typical" curve relating the ratio of peak average
acceleration to peak ground acceleration to the depth of the failure surface for earth dams founded
on rock. Kavazanjian and Matasovi (1995) demonstrated that the Makdisi and Seed (1978) earth
dams curve provides a reasonable representation of the profile of average acceleration versus
depth in solid waste landfills over 50 ft (15 m) thick. Figure 4.7 (Kavazanjian and Matasovi,
1995) compares nine different solid waste landfill non-linear seismic response analyses,
encompassing waste fills from 50 to 300 ft (15 to 90 m) thick to the representative profile
developed by Makdisi and Seed. Based upon a maximum average acceleration ratio at the base
of the landfill of 0.45, as indicated by Figure 4.7, and upon a maximum amplification factor of
2.0 from Figure 4.5 for a peak bedrock acceleration of 0.1 g or greater, Kavazanjian and
Matasovi (1995) concluded that the free field peak ground acceleration calculated for the landfill
46
-------
site in Step 2 provides a conservative estimate of the peak average acceleration at the base of the
landfill for use in analyses of base liner stability and seismic deformation potential.
Bray, et al. (1995) provide a means of quantifying the limitations of the Kavazanjian and Matasovi
(1995) caveat of fill thickness and of more precisely evaluating the average acceleration of the
waste mass as a function of the peak ground acceleration. Figure 4.8, from Bray, et al. (1995),
presents a normalized plot of the peak acceleration at the crest of the landfill divided by the peak
average acceleration of the waste mass versus the ratio of the fundamental period of the waste
mass to the fundamental period of the design earthquake from results of a large number of landfill
response analyses with peak ground accelerations up to 0.35 g. The fundamental period of the
waste fill is the reciprocal of the fundamental frequency evaluated using Equation 4.1.
Based upon the mean plus two standard deviation curve, Figure 4.8 indicates that the peak average
acceleration of the waste mass is equal to or less than the peak ground acceleration when the
fundamental period of the waste mass is at least 1.2 times greater than the fundamental period of
the design earthquake. Based upon typical shear wave velocities for solid waste and typical
predominant periods for earthquake motions, Figure 4.8 suggest that the peak average acceleration
of the waste mass can be assumed to be less than the free field peak ground acceleration for
nearby earthquakes for waste thicknesses greater than 50 ft (15 m) and for distant earthquakes for
waste thicknesses greater than 100 ft (30 m). For larger waste thicknesses and/or high frequency
(short period) earthquakes, Figure 4.8 indicates the peak average acceleration of the waste mass
can be as little as 20 to 40 percent of the free field peak ground acceleration.
4.1.2 One-Dimensional Site Response Analysis
For Special Study soil sites, for major projects, and when an analysis more accurate than the
simplified one presented in the previous section is desired, a one-dimensional seismic site response
analysis can be performed. The site response analysis can be performed for the foundation soils
only, for the waste mass only, or for the coupled response of the foundation soil and waste mass,
depending on the needs and desires of the design engineer.
The computer program, SHAKE, originally developed by Seed and his co-workers (Schnabel et
al., 1972) and recently updated by Idriss and Sun (1992) is perhaps the most commonly used
computer program for one-dimensional equivalent linear seismic site response analysis. Basic
input to SHAKE includes the soil profile, soil properties, and the input time history. Soil
properties include the maximum (small strain) shear wave velocity or shear modulus and unit
weight for each soil layer plus curves relating the reduction in modulus and damping ratio to shear
strain for each soil type.
47
-------
Modulus reduction and damping curves can be specified by the user based upon laboratory testing
or upon recommendations from previous investigations. Laboratory data on soil modulus and
damping at small strains (shear strains less than 10"4%) can be obtained from resonant column
tests. At larger strains, cyclic simple shear, cyclic triaxial, and cyclic torsional shear tests can
be used. American Society for Testing and Materials (ASTM) standards exists for resonant
column testing (ASTM D-3999) and cyclic triaxial testing (ASTM D-4015). Small strain modulus
can also be determined from field measurements of shear wave velocity. Shear wave velocity can
be measured in the field using geophysical methods such as down-hole and cross-hole velocity
testing, seismic refraction, and spectral analyses of surface waves. Field measurements are
generally considered more reliable than laboratory measurements of shear wave velocity or small
strain modulus. Field techniques for measurement of the dynamic modulus at large strains and
of the damping ratio are not currently available. Shear wave velocity is related to small strain
shear modulus, Gmax, by the equation:
(4-2)
As an alternative to laboratory or field measurement of soil properties, dynamic moduli and
damping for soils may be estimated as a function of soil type based upon recommendations for
typical values from previous investigations. One set of practical recommendations for estimating
modulus and damping of typical soils are summarized in Figure 4.9 and Table 4.1 Figure 4.9
presents typical modulus reduction and damping curves as a function of the plasticity index of the
soil, PI, from Vucetic and Dobry (1991). (Note that the curve for PI = 0 represents sands and
cohensionless soils.) These curves are for all soil types for a broad range of overconsolidation
ratios. Table 4.1 presents coefficients and exponents for evaluating the small strain shear modulus
for different soil types using the Standard Penetration Test blow count, N, and the following
equation from Imai and Tonouchi (1982):
G,nax W (4-3)
where N is in blows per foot of penetration and c and a are coefficients from Table 4.1. Equation
4.3 was developed using Japanese data. Therefore, a blowcount corresponding to hammer
efficiency of 60 percent, N^, as used in U.S. practice (described in Section 5.3), needs to be
converted to Japanese standards by multiplying NM by 0.833 before input to Equation 4.3:
N 0.833(tf60) (4.4)
48
-------
Other correlations for these properties are available and may be used. Due to the uncertainty
involved in using these types of empirical correlations, considerable judgement is required in
interpreting results of analyses that employ them and sensitivity studies on the values of these
parameters are recommended.
Unit weight, shear modulus, and damping values are also required for MSW if the MSW is
included in the response analysis. Measurement of the dynamic properties of MSW in the
laboratory is considered neither practical nor reliable due to the difficulties inherent to sampling
and testing MSW. Back calculation of MSW properties from field observations is generally
considered to be the most reliable means of evaluating these properties at this time (Kavazanjian
et al., 1995). Evaluation of the density of MSW from reported field measurements is discussed
in Section 6.1.
At present, the shear wave velocity of MSW has been measured in-situ at a limited number of
locations. Cross-hole shear wave velocity measurements at the Puente Hills MSW Landfill in
southern California reported by Earth Technology (1988) varied from 213 m/s (700 ft/s) at the
ground surface to 278 m/s (920 ft/s) at a depth of approximately 14 meters (45 feet). Sharma et
al. (1990) report an average shear wave velocity of 198 m/s (455 ft/s) for MSW at depths between
0 and 15 meters (0 and 50 feet) at a landfill in Richmond, California from downhole shear wave
velocity measurements. Singh and Murphy (1990) cite an investigation by others at the Redwood
Landfill in the San Francisco Bay area where an average shear wave velocity of 91 m/s (300 ft/s)
was reported for the refuse. Shear wave velocities backfigured using assumed values of Poisson's
ratio and waste density from Young's Modulus values developed by Carey et al. (1993) from
cross-hole shear wave velocity measurements vary from 185 m/s (610 ft/s) near the surface to 478
m/s (1,580 ft/s) at a depth of 30 meters (100 feet) at the Brookhaven landfill on Long Island in
New York (actual shear wave velocity measurements were not reported). Measurements at 8
MSW landfills in southern California made using Spectral Analysis of Surface Waves (SASW)
were reported by Kavazanjian et al. (1994). Shear wave velocities varied from 78 to 170 m/s (260
to 560 ft/s) near the ground surface, and from 150 to 300 m/s (500 to 990 ft/s) at a depth of 20
meters (66 feet). Shear wave velocity was reported to increase steadily with depth in the waste
at all 8 sites.
Hushmand Associates (1994) report that seismic refraction surveys performed by others at the Oil
landfill yielded a shear wave velocity of between 200 to 240 meters per second (660 to 800 feet
per second). Hushmand Associates (1994) also report that measurements of micro tremors from
small earthquakes and of ambient vibrations at a strong motion instrumentation station located
over an estimated 75 meters (250 feet) of waste at the Oil site indicate a predominant period of
between 0.8 and 1.2 seconds (corresponding to a predominant frequency of between 1.25 and 0.83
49
-------
cycles per second) for the waste mass. Using Equation 4.1, this corresponds to an average shear
wave velocity of between 240 and 360 meters per second (800 and 1200 feet per second) for the
assumed 75 meter (250 foot) waste column. While the Oil landfill is composed of mixed
industrial and municipal waste, the portion of the landfill at which the strong motion station is
located is believed to be composed primarily of MSW (personnel communication, Professor R.B.
Seed, U.C. Berkeley, to Dr. Edward Kavazanjian, Jr., GeoSyntec Consultants).
Based upon the data cited above, Kavazanjian et al. (1995) developed a "representative" shear
wave velocity profile for MSW landfills. Figure 4.10 (Kavazanjian et al., 1995) presents a
composite plot of the available MSW shear wave velocity data along with the shear wave velocity
profile developed by these investigators for use in the absence of site-specific data. In developing
this shear wave velocity profile, the seismic refraction data from the Oil site was assumed to
represent the average velocity over the top 30 meters (100 feet) of waste and the data derived from
cross-hole measurements was considered unreliable due to the potential for "short-circuiting" of
the wave travel path by layers of daily and intermediate cover soils.
Modulus reduction and damping curves for MSW have never been measured in the laboratory.
Prior to the 17 January 1994 Northridge earthquake, no data was available to back-calculate MSW
modulus and damping from the observed seismic response of landfills. In the absence of special
measurements, most investigators based the modulus reduction and damping curves for MSW
upon those of clay and peat soils (Earth Technology, 1988; Singh and Murphy, 1990; Sharma and
Goyal, 1991; and Repetto et al., 1993). Figure 4.11 presents recommendations from Earth
Technology (1988) for modulus reduction and damping curves for MSW. These curves are
reported to be based upon modulus reduction curves for peat and damping curves for clay. Figure
4.12 presents recommendations for modulus reduction and damping in MSW from Singh and
Murphy (1990). The "recommended" curves are described by Singh and Murphy as the "average"
of typical modulus reduction and damping curves for peat and clay that are used in engineering
practice.
The strong motion recordings captured at the Oil landfill in the M 6.7 Northridge earthquake
represent the first (and currently the only) direct measurement of the seismic response of a solid
waste landfill. In the Northridge event, the peak ground acceleration at the monitoring station on
the rock outcrop adjacent to the landfill was 0.25 g, while the peak ground acceleration at the top
of the landfill was 0.24 g (Hushmand Associates , 1994). Time histories of acceleration, velocity,
and displacement recorded at the top of the landfill for one horizontal component of motion were
previously presented in Figure 3.4.
Kavazanjian and Matasovi (1995) developed the MSW modulus reduction and damping curves
shown in Figure 4.13 from the observed response of the Oil landfill in the Northridge event.
50
-------
Using the representative shear wave velocity profile shown in Figure 4.8 and the "typical" unit
weight profile developed by Kavazanjian et al. (1995), presented subsequently in Figure 6.3 of
this document, Kavazanjian and Matasovi (1995) back-calculated parameters describing the cyclic
behavior of MSW for a non-linear site response model from the observed landfill response. Then,
these investigators used the non-linear model to predict the response of MSW to uniform cyclic
loading and compute the modulus reduction and damping curves for MSW shown in Figure 4.13.
The modulus reduction and damping curves for MSW shown in Figure 4.13 were used by
Kavazanjian and Matasovi (1995) in the program SHAKE to perform an equivalent linear response
analysis of the response of the Oil landfill in the Northridge earthquake. By trial and error, a
representative shear strain factor of 0.8 (representative shear strain equal to 0.8 times the
maximum cyclic shear strain) was found to give the best agreement between observed and
predicted behavior.
Figure 4.14 compares the Oil landfill response observed in the Northridge earthquake to the
response predicted by Kavazanjian and Matasovi (1995) using SHAKE, the modulus reduction and
damping curves in Figure 4.12, and a representative shear strain factor of 0.8. This figure also
shows landfill response predicted by Kavazanjian et al. (1995) using SHAKE and various
combinations of modulus reduction and damping curves for peat and clay along with the best fit
representative shear strain factor. Based upon this comparison, Kavazanjian et al. (1995)
suggested that the modulus reduction and damping curves shown in Figure 4.13 be used in
equivalent linear seismic response analysis of MSW landfills until additional information on the
cyclic response of MSW becomes available.
The use of a representative shear strain factor of 0.8 with the Kavazanjian and Matasovi modulus
reduction curve indicates that the Oil landfill behaved relatively elastically in the Northridge
earthquake. Furthermore, the maximum cyclic shear strain induced in the Oil landfill during the
Northridge event was on the order of 2 x 102 percent. Therefore, the shape of the modulus
reduction and damping curves shown in Figure 4.13 must be considered speculative for values of
shear strain greater than 2 x 10~2 percent. Furthermore, there is some concern over geological
conditions at the location where the base motions of the Oil landfill were recorded. For these
reasons, caution is warranted in using the modulus reduction and damping curves presented in
figure 4.13.
4.1.3 Two- and Three-Dimensional Site Response Analysis
Computer programs are available for equivalent linear and truly non-linear two- and three-
dimensional seismic site response analyses. However, such programs are not commonly used in
landfill engineering practice. The programs for two- and three-dimensional site response analyses
51
-------
are not particularly "user-friendly". Furthermore, experience with two-dimensional site response
analyses of earth dams has shown that one-dimensional site response analyses of vertical columns
within the embankment typically yield accelerations and stresses within ten percent of the results
of the more sophisticated two- and three-dimensional analyses (Vrymoed and Calzascia, 1978).
Two- and three-dimensional effects may logically be expected to be even less significant with
respect to the seismic response of landfills compared to earth dams, as landfills tend to be massive
structures with broad decks. Considering the level of uncertainty associated with material
properties of solid waste, two- and tree-dimensional seismic response analyses do not appear to
be warranted for most municipal solid waste landfill projects at this time.
Once the soil profile and material properties have been specified, the only remaining input is the
input earthquake motion. Selection of representative time histories for the input motion is
discussed in Section 4.2.
4.2 Selection of Earthquake Time History
Earthquake time histories may be required for input to SHAKE seismic response analyses or, if
a simplified seismic response analysis is employed, for input to the seismic deformation analyses
described in Section 6. Time histories can be developed either by selecting a representative time
history from the catalog of acceleration time histories recorded in previous earthquakes or by
synthesizing an artificial accelerogram. Time histories should be developed for each significant
souce impacting the site.
Selection of a representative time history from the catalog of available strong motion records and
scaling it to the appropriate peak acceleration is, in general, a preferable approach to use of a
synthetic time history. However, due to limitations in the catalog of available records, it is not
always possible to find a representative time history from the catalog of available records,
particularly for the eastern and central United States.
In selecting a representative time history from the catalog of available records, an attempt should
be made to match as many of the relevant characteristics of the design earthquake as possible.
Important characteristics that should be considered in selecting a time history from the catalog
include:
earthquake magnitude;
source mechanism (e.g., strike-slip, normal, or reverse faulting);
focal depth;
site to source distance;
site geology; and
52
-------
peak ground acceleration.
These factors are ranked in a general order of decreasing importance. However, the relative
importance may vary from case to case. For instance, if a bedrock record is chosen for use in a
SHAKE analysis of the influence of local soil conditions, site geology will not be particularly
important in selection of the input bedrock time history. However, if a soil site record scale to
a peak ground acceleration that already includes consideration of the potential for amplification
of motions by local soil conditions is to be used in the response analysis, site geology can be a
critical factor in selection of an appropriate time history.
Scaling of the peak acceleration of a record by a factor of more than two is not recommended,
as the frequency characteristics of ground motions can be directly and indirectly related to the
amplitude of the motion. Leeds (1992) and Naeim and Anderson (1993) present summaries of
available strong motion records and their characteristics.
Due to uncertainties in the selection of a representative earthquake time history, response analyses
should never be performed using only a single time history. The use of a suite time histories is
recommended for purposes of evaluating seismic site response. Engineers commonly use three
and sometimes as many as five time histories to represent each significant seismic source in a site
response analysis. For earthquakes in the western United States, it should be possible to find
three to five representative time histories that satisfy the above criteria. However, at the present
time, there are only two bedrock strong motion records available from earthquakes of magnitude
M 5.0 or greater in the central and eastern North America:
the Les Eboutements record with a peak horizontal acceleration of 0.23 g from
the 1988 Saguenay, Quebec earthquake of magnitude M 6.0; and
the Loggie Lodge record with a peak horizontal acceleration of 0.4 g from the
1981 Mirimichi, New Brunswick earthquake of magnitude M 5.0.
Therefore, for analysis of sites east of the Rocky Mountains, at least one record from a western
United States site, an international recording site, or a synthetic accelerogram is required to
compile a suite of three records for analysis. For the new Madrid seismic zone, where neither
the Mirimichi nor Saguenay record is of appropriate magnitude, all three records must be from
either the western United States, an international site, or synthetically generated.
One of the primary differences anticipated between earthquakes in the eastern and central United
States and those in the western United States is frequency content (Nuttli, 1981; Atkinson, 1987).
There may also be a difference in duration due to the different rates of acceleration attenuation.
53
-------
For liquefaction analyses which depend only on peak acceleration, use of a western United States
earthquake record of appropriate magnitude and intensity for analysis of a site in the eastern or
central United States should be acceptable. However, for analysis of seismic deformation
potential at an eastern or central United States site, the appropriateness of using a western United
States earthquake record is uncertain. The greater energy at lower frequencies in typical western
U.S. records could result in a conservative estimate of deformation potential at an eastern or
central United States site. On the other hand, the potential for a longer duration on the east coast
compared to the west coast for an earthquake of the same magnitude and distance could have the
opposite effect.
Due to the difference in the anticipated depths of the causative faults, when using a western United
States record to analyze a site in the eastern United States precedence should be given to matching
hypocentral distance over peak acceleration. Hypocentral distance is the distance from the site
to the center of energy release for the earthquake. Hypocentral distance includes the effect of the
depth of the earthquake in the distance measure.
Computer programs are available to generate a synthetic seismic accelerogram to meet peak
acceleration, duration, and frequency content requirements (Gasparin and Vanmarcke, 1976; Ruiz
and Penzien, 1969; Silva and Lee, 1987). Synthetic earthquake accelerograms for many regions
of the country are currently being compiled by Dr. Klaus Jacob at the Lamont-Doherty
Observatory of Columbia University under the auspices of the National Center for Earthquake
Engineering Research (see Table 2.2). However, at the time of preparation of this guidance
document this compilation was not yet available. The generation of synthetic acceleration time-
histories is not generally within the technical expertise of civil engineering firms and should not
be undertaken without expert consultation. For this reason, generation of synthetic earthquake
acceleration time-histories is beyond the scope of this manual. However, appropriate synthetic
accelerograms may be available to the engineer from previous studies and may be used if they are
shown to be appropriate for the site.
Comparison of acceleration of response spectra from candidate accelerograms to acceleration
response spectra deemed representative of the design event provides useful means of determining
whether the selected accelerograms are indeed representative. Each accelerogram from the
selected suite of accelerograms should fall primarily within the two-sigma boundaries of the
statistically-derived response spectra and the suite of accelerogram should average out to close to
the near spectra. In this manner, a representative suite of time histories can be developed.
54
-------
4.3 References
Anderson, D.G., Hushmand, B., and Martin, G.R. (1992), "Seismic Response of Landfill Slopes,
" Proc. Stability and Performance of Slopes and Embankments - II, Vol. 2, ASCE Geotechnical
Special Publication No. 31, Berkeley, California, pp. 973-989.
Atkinson, G.M. (1987), "Implications of Eastern Ground Motion Characteristics for Seismic
Hazard Assessment in Eastern North America," Proc. Symposium on Seismic Hazards, Ground
Motions, Soil-Liquefaction and Engineering Practice in Eastern North America, Tuxedo, New
York, NCEER Technical Report No. NCEER-87-0025.
Borcherdt, R.D. (1994) "New Developments in Estimating Site Effects on Ground Motion," Proc.
Seminar on New Developments in Earthquake Ground Motion Estimation and Implications for
Engineering Design Practice, Applied Technology Council, ATC35-1, Redwood City, California,
pp. 2-1 - 2-12.
Bray, J.D., Augello, A.J., Leonards, G.A., Repetto, P.C., and Byrne, RJ. (1995), "Seismic
Stability Procedures for Solid-Waste Landfills," Journal of the Geotechnical Division, ASCE,
Vol. 121, No. 2, pp. 139-151.
Carey, P.J., Koragappa, N. and Gurda, J.J. (1993) "A Case Study of the Brookhaven Landfill
-Long Island, New York," Proc. Waste Tech '93, Marina del Rey, California, 15 p.
Earth Technology (1988), "In-Place Stability of Landfill Slopes, Puente Hills Landfill, Los
Angeles, California," Report No. 88-614-1, The Earth Technology Corporation, Long Beach,
California.
Gasparin, D.A. and Vanmarcke, E.H (1976) "SIMQKE - A Program for Artificial Motion
Generation," Department of Civil Engineering, Massachusetts Institute of Technology,
Massachusetts.
GeoSyntec (1994), "Non-Linear Seismic Response Analysis of Solid Waste Landfills," Internal
Research Report, GeoSyntec Consultants, Huntington Beach, California.
Harder, L.S., Jr. (1991), "Performance of Earth Dams During the Loma Prieta Earthquake,"
Proceedings of the Second International Conference on Recent Advances in Geotechnical
Earthquake Engineering and Soil Dynamics, University of Missouri, Rolla, 11-15 March.
55
-------
Hushmand Associates (1994), "Landfill Response to Seismic Events," Report prepared for the
USEPA Region IX, Hushmand Associates, Laguna Niguel, California.
Idriss, I.M. (1990), "Response of Soft Soil Sites During Earthquakes," Proc. Symposium to Honor
ProfessorH.B. Seed, Berkeley, California.
Idriss, I.M. and Sun, J.I. (1992) "User's Manual for SHAKE91," Center for Geotechnical
Modeling, Department of Civil and Environmental Engineering, University of California, Davis,
California, 13 p. (plus Appendices).
Imai, T. and Tonouchi, K. (1982), "Correlation of N Value with S-Wave Velocity," Proc. 2nd
European Symposium on Penetration Testing, Amsterdam, The Netherlands, pp. 67-72.
Kavazanjian, E., Jr., and Matasovi, N. (1995), "Seismic Analysis of Solid Waste Landfills,"
Proceedings of the Geoenvironment 2000 Specialty Conference, ASCE, Vol. 2, pp. 1066-1080,
New Orleans, Louisiana, 24-26 February .
Kavazanjian, E., Jr., Snow, M.S., Matasovi, N., Poran, C., and Satoh, T. (1994a), "Non-
Intrusive Rayleigh Wave Investigations at Solid Waste Landfills." Proc. 1st International
Congress on Environmental Geotechnics, Edmonton, Alberta, pp. 707-712.
Kavazanjian, E., Jr., Matasovi, N., Bonaparte, R., and Schmertmann, G.R. (1995), "Evaluation
of MSW Properties for Seismic Analysis," Proceedings of the Geoenvironment 2000 Specialty
Conference, ASCE, Vol. 2, pp. 1126-1141, New Orleans, Louisiana, 24-26 February .
Lee, M.K.W., and Finn, W.D.L. (1978), "DESRA-2, Dynamic Effective Stress Response
Analysis of Soil Deposits with Energy Transmitting Boundary Including Assessment of
Liquefaction Potential," Soil Mechanics Series No. 36, Department of Civil Engineering,
University of British Columbia, Vancouver, Canada, 60 p.
Leeds, D.L. (1992) "State-of-me-Art for Assessing Earthquake Hazards in the United States:
Report 28, Recommended Accelerograms for Earthquake Ground Motions," Misc. Paper S-73-1,
Geotechnical Laboratory, U.S Army Waterways Experiment Station, Vicksburg, Mississippi, 171
p. (plus Appendices).
Matasovi, N. and Vucetic, M. (1993), "Cyclic Characterization of Liquefiable Sands," Journal
of Geotechnical Engineering, ASCE, Vol. 119, No. 11, pp. 1805-1822.
56
-------
Makdisi, F.I. and Seed, H.B. (1978), "Simplified Procedure for Estimating Dam and
Embankment Earthquake-Induced Deformations," Journal of the Geotechnical Division, ASCE
Vol. 104, No. 4 , pp. 849-867.
Naeim, F. and Anderson, J.C. (1993), "Classification and Evaluation of Earthquake Records for
Design," Report No. CE 93-08, Department of Civil Engineering, University of Southern
California, Los Angeles, 288 p.
Nuttli, O.W. (1981), "Similarities and Differences Between Western and Eastern United States
Earthquakes, and their Consequences for Earthquake Engineering," Earthquakes and Earthquake
Engineering: The Eastern United States, Vol. 1, Assessing the Hazard - Evaluating the Risk, I.E.
Beavers, Ed., Ann Arbor Science Publishers, Inc., Ann Arbor, Michigan, pp. 25-51.
Repetto, P.C., Bray, J.D., Byrne, R.J. and Augello, A.J., (1993), "Applicability of Wave
Propagation Methods to the Seismic Analysis of Landfills," Proc. Waste Tech '93, Marina Del
Rey, California, pp. 1.50-1.74.
Ruiz, J. and Penzien, J., (1969) "PSEQGN - Artificial Generation of Earthquake Accelerograms,"
Report No. EERC 69-3, Earthquake Engineering Research Center, University of California,
Berkeley, California.
Schnabel, P.B., Lysmer, J. and Seed, H.B. (1972) "SHAKE: A Computer Program for
Earthquake Response Analysis of Horizontally Layered Sites." Report No. EERC 72-12,
Earthquake Engineering Research Center, University of California, Berkeley, California.
Seed, H.B. (1975) "Earthquake Effects on Soil-Foundation Systems," In Foundation Engineering
Handbook, H.F. Winterkorn and H.Y. Fang Eds., Van Nostrand Reinhold, New York,
pp. 700-732.
Seed, H.B. and Idriss, I.M., (1982), "Ground Motions and Soil Liquefaction During
Earthquakes," Monograph No. 5, Earthquake Engineering Research Institute, Berkeley,
California, 134 p.
Sharma, H.D., Dukes, M.T., and Olsen, D.M., (1990), "Field Measurements of Dynamic Moduli
and Poisson's Ratio of Refuse and Underlying Soils at a Landfill Site," In Geotechnics of Waste
Fill - Theory and Practice, ASTM STP 1070, pp. 57-70.
57
-------
Sharma, H.D. and Goyal, H.K. (1991), "Performance of a Hazardous Waste and Sanitary Landfill
Subjected to Loma Prieta Earthquake," Proc. 2nd International Conference on Recent Advances
in Geotechnical Earthquake Engineering and Soil Dynamics, St. Louis, Missouri, pp. 1717-1725.
Silva, W.D. and Lee, K. (1987) "State-of-the-Art for Assessing Earthquake Hazards in the United
States: Report 24, WES RASCAL Code for Synthesizing Earthquake Ground Motions," Misc.
Paper S-73-1, Geotechnical Laboratory, U.S. Army Waterways Experiment Station, Vicksburg,
Mississippi.
Singh, S. and Murphy, B.J., (1990), "Evaluation of the Stability of Sanitary Landfills," In
Geotechnics of Waste Fills - Theory and Practice, ASTM STP 1070, pp. 240-258.
Singh, S. and Sun, J.I. (1995), "Seismic Evaluation of Municipal Solid Waste Landfills,"
Proceedings of the Geoenvironment 2000 Specialty Conference, ASCE, Vol. 2, pp. 1081-1096,
New Orleans, Louisiana, 24-26 February.
Vrymoed, J.L. and Calzascia, E.R. (1978), "Simplified Determination of Dynamic Stresses in
Earth Dams," Proc. Earthquake Engineering and Soil Dynamics, ASCE, Pasadena, California,
pp. 991-1006.
Vucetic, M. and Dobry, R. (1991) "Effect of Soil Plasticity on Cyclic Response." Journal of the
Geotechnical Engineering, ASCE, Vol. 117, No. 1, 89-107.
Wood, H.O. (1908), "Distribution of Apparent Intensity in San Francisco," in The California
Earthquake of April 18, 1906," Report of the State Earthquake Investigation Commission,
Carnegie Institution of Washington, D.C., pp. 220-245.
58
-------
TABLE 4.1: PARAMETERS FOR THE EMPIRICAL RELATIONSHIP
TO ESTIMATE G^
(After Imai and Tonouchi, 1982)
SOIL TYPE
Peat
Clay
Sand
Gravel
c
(kg/cm2)
53.7
176.0
125.0
82.5
a
(-)
1.08
0.607
0.611
0.767
Notes: (1) G,^ = Small strain shear modulus; G^ = c(Nf; G^ in kg/cm2.
(2) N ~ (uncorrected) SPT blowcount according to Japanese standards. Multiply Nw from
U.S. practice by 0.833 to estimate a comparable blow count.
(3) Correlation applies only for soils of alluvial origin. For soils of other origin, the original
reference should be consulted.
59
-------
0 0.5 1.0
Perted-Secendi
06
0.6
• 0.4
0.2
0
0.5
04
03
02
01
0,
0 0.5 1.0152.00 OS 1.0
03 IB 15 20
n
'\
V
v^j [__
A
*s,
A,
A
Vp
-^
-
Period-Seconds
Perted-l
Perlod-S
\SZQ 0 05 15 1520
MSfkxJ* Seconds
Figure 4.1 Soil Conditions and Characteristics of Recorded Ground Motions, San
Francisco M 5.7 Earthquake of 22 March 1957 (Seed, 1975).
-------
ACCELERATION
RESPONSE
SPECTRA
PEAK GROUND ACCELERATION
TIME IN
SECONDS
SYSTEM
RESPONSE
ATT.
PERIOD IN
SECONDS
Figure 4.2 Development of Acceleration Response Spectrum for Damped Single
Degree of Freedom System.
61
-------
40O
200 -
1OO
Ul
CO
CO
s
o
o
UJ
.CM .06
.1 3. .4 .6 1 2 4 6 10
PERIOD, SECONDS
SflN FERNHNOO EflRTHOURKE FEB 9. 1971 - 0600 PST
IllCOta 71.008.O 6MH ORION ».VO. 1ST fLOOft. LOS BKCLtS. CSL. COW WOW
VflLUtS IWt 0. 3. 5. 10 BNO ?0 PCKTCNT (f CR1I1CRL
Figure 4.3 Tripartite Representation of Response Spectra.
62
-------
o
a
o
x
«j
0.6
0.5
O.-4
Soft to Modknti
Stiff Clay and Sand
.1 O.2 O.3 0.4 O.5 O.6
Maximum Acceleration in Rock (g)
O.7
O.t 0.2 0.3 0.4 0.5
Acceleration at Rock Sites - g
0.6
Figure 4.4 Relationship Between Maximum Acceleration on Rock and Other Local Site
Conditions: (a) Seed and Idriss (1982); (b) Idriss (1990).
63.
-------
0.8
to
•««*'
cd
1
0.7 -
0.6 -
0.0
A OH Landfill - Recorded
•^Landfills — Non—linear Analyses
Soft Ground Sites — Recorded
Idriss (1990) - Recommended
1979 Montenegro Eq.
1989 Loma Prieta Eq.
1985 Mexico City Eq.
i i i i | i i i i | i
0.0 0.1 0.2
i i i i i i i j i i i i j r
0.3 0.4 0.5
i
PEAK OUTCROP ACCELERATION
a
omax
0.6
(g)
Figure 4.5 Observed Variations of Peak Horizontal Accelerations on Soft Soil and
MSW Sites in Comparison to Rock Sites (Kavazanjian and Matasovic,
1994).
64
-------
o>
Hardar (1991)
Upper Bound for Observed
Motions at Earth Oaras
Seed - Idriss (1982) for
Deep Coftesiontess Soil
0.2 0.3 0.4 0.5
Max. Acd. at Base, g
0.6 0.7
Figure 4.6 Approximate Relationship Between Maximum Accelerations at the Base
and Crest for Various Ground Conditions (Singh and Sun, 1995)
65
-------
Average Acceleration Ratio, a
0.0 0.2 0.4 0.6 0.8
0.0
0.2
D-MOD (3
Haltdiai tc Sect
(z=0)
1.0
Figure 4.7 Variation of Maximum Average Acceleration Ratio with Depth of Sliding
Mass (Kavazanjian and Matasovic, 1995)
66
-------
^ 3
2.5
t 2
s
4 1
o
0.5
0 mean +2 std. dev.
o
mean -2 std. dev.
246
Normalized Fundamental Period (Tw/Tp-eq)
8
Figure 4.8 Normalized Maximum Horizontal Equivalent Acceleration versus the
Normalized Predominant Period of the Waste Fill (Bray, et al., 1995)
67
-------
1.0
0.0001
0.001 0.01 0.1 1
CYCLIC SHEAR STRAIN, 7C(%)
0.0001
0.001 0.01 0.1 1
CYCLIC SHEAR STRAIN, 7C (%)
10
Figure 4.9 Modulus Reduction and Damping Curves for Soils of Different Plasticity
Index (PI) (Vucetic and Dobry, 1991).
68
-------
0
30-
120
This Study
Kavazanjian et al. (1994) (8 sites) \
after Carey et al. (1993) \
Shanna et al. (1990)
Woodward-Clyde Consultants (1987)
after Hushmand Associates (1994)
after Earth Technology (1988) (crosshole)
after Earth Technology (1988) (downhole)
0
1 I I
100
300
200 300 400 500
SHEAR WAVE VELOCITY (m/s)
600
Figure 4.10 Shear Wave Velocity of MSW (Kavazanjian et al., 1995)
69
-------
0.8
0.6
I
a
0.4
0.2
0
io-*
10-3
-PEAT (SEED AND IORISS. 1970).
MODIFIED FOR THIS PROJECT
10-2
SHEARING STRAIN <%)
CLAY (SEED AND IORISS. 1970
10 -1
SHEARING STRAIN (%)
100
10
Figure 4.11 Modulus Reduction and Damping Curves for MSW (Earth Technology,
1988).
70
-------
3000
CO
\
CD
PRA(1987)
Upper/Lower
'Bound tor SHAKE.
Landfill Refuse Material
Assumed to Correspond to the
Statistical Average ot Peat & Clay
ech (1988)
Poirrte-Hill Fill
Redwood Fill
Nota: Oala Points are estimated average
•aloes torn shear wave data tor Su-143 KNAn*
I 1 i I
0.001 0.01 0.1 1.0
CYCLIC SHEAR STRAIN IN %
T
10
20-
O 15-
P:
CD 10-
^
S;
1 5-
Landfill Refuse Material
Assumed to Correspond to the
Statistical Average of Peat & Clay
1——— 1 r
0.001 0.01 0.1 1.0
CYCLIC SHEAR STRAIN IN %
10
Figure 4.12 Modulus Reduction and Damping Curves for MSW (Singh and Murphy,
1990).
71
-------
§0.00
0.0001 0.001 0.01
CYCLIC SHEAR STRAIN
0.0001 0.001 0.01 0.1 1
CYCLIC SHEAR STRAIN (SB)
Figure 4.13 Modulus Reduction and Damping Curves for MSW (Kavazanjian and
Matasovic, 1995).
72
-------
bfll-2
^«_x
§1.0
0.8
0.6
O
05
H
O
L4 -
0.2 -
53 o.o
RECORDED (On-Northxidge; Longitud.)
MSW (Kavazanjian tt Matasovic, 1995)
CIAY (PI=15) (Vucetic & Dobry. 1991)
CLAY (Seed & Idrias, 1970b)
PEAT (Seed & Idriss. 1970a)
Damping =
0.0
I I | i
0.5
1.0 1.5 2.0 2.5
PERIOD (sec)
3.0 3.5 4.0
Figure 4.14 Comparison of Oil Landfill Response to Results of Equivalent Linear
Analysis (Kavazanjian et al., 1995).
-------
SECTION 5
258.14 SEISMIC IMPACT ZONES:
LIQUEFACTION ANALYSIS
During strong earthquake shaking, loose, saturated conesionless soil deposits may experience a
sudden loss of strength and stiffness, sometimes resulting in large, permanent displacements of
the ground. This phenomenon is called soil liquefaction. Liquefaction beneath and in the vicinity
of a municipal solid waste landfill facility (MSWLF) can have severe consequences with respect
to the integrity of the landfill containment system. Localized bearing capacity failures, lateral
spreading, and excessive settlements resulting from liquefaction may damage landfill liner and
cover systems. Liquefaction-associated lateral spreading and flow failures can also affect the
global stability of the landfill. Therefore, a liquefaction potential assessment is a key element in
the seismic design of landfills.
This Section outlines the current state-of-the practice for evaluation of the potential for soil
liquefaction and the consequences of soil liquefaction (should it occur) as it applies to the seismic
design of a MSWLF. Initial screening criteria to determine whether or not a liquefaction analysis
is needed are presented in Section 5.1. The simplified procedure for liquefaction potential
assessment commonly used in engineering practice is presented in Section 5.2. Methods for
performing a liquefaction impact assessment are presented in Section 5.3. Methods for mitigation
of liquefaction potential and the consequences of liquefaction are discussed in Section 5.4.
Advanced methods for liquefaction potential assessments, including one- and two-dimensional
fully-coupled effective stress site response analyses, are also discussed in Section 5.4.
5.1 Initial Screening
The first step in any liquefaction evaluation is to assess whether the potential for liquefaction of
cohesionless soils exists at a site. A variety of screening techniques exists to distinguish sites that
are clearly safe with respect to liquefaction from those sites that require more detailed study (e.g.,
Dobry et al., 1980). The following five screening criteria are most commonly used to make this
assessment:
Geologic age and origin. Liquefaction potential decreases with increasing age of
a soil deposit. Pre-Holocene age soil deposits generally do not liquefy, though
liquefaction has occasionally been observed in Pleistocene-age deposits. Table 5.1
presents the liquefaction susceptibility of soil deposits as a function of age and
origin (Youd and Perkins, 1978).
74
-------
Fines content and plasticity index. Liquefaction potential decreases with increasing
fines content and increasing plasticity index, PI. Data presented in Figure 5.1
(Ishihara et al., 1989) show grain size distribution curves of soils known to have
liquefied in the past. This data serves as a rough guide for liquefaction potential
assessment of cohesionless soils. Soils having greater than 15 percent (by weight)
finer than 0.005 mm, a liquid limit greater than 35 percent, and an in-situ water
content less than 0.9 times the liquid limit generally do not liquefy (Seed and
Idriss, 1982).
Saturation. Although partially saturated soils have been reported to liquefy, at
least 80 to 85 percent saturation is generally deemed to be a necessary condition
for soil liquefaction. In many locations, the water table is subject to seasonal
oscillation. In general, it is prudent that the highest anticipated seasonal water
table elevation be considered for initial screening.
Depth below ground surface. While failures due to liquefaction of end-bearing
piles resting on sand layers up to 100 ft (30 m) below the ground surface have been
reported, surface effects from liquefaction is generally not likely to occur more
than 50 ft (15 m) below the ground surface.
Soil Penetration Resistance. According to the data presented in Seed and Idriss
(1985), liquefaction has not been observed in soil deposits having normalized
Standard Penetration Test (SPT) blowcount, (N^ larger than 22. Marcuson, et
al. (1990) suggest a normalized SPT value of 30 as the threshold value above
which liquefaction will not occur. However, Chinese experience, as quoted in
Seed et al. (1983), suggests that in extreme conditions liquefaction is possible in
soils having normalized SPT blowcounts as high as 40. Shibata and Teparaska
(1988), based on a large number of observations, conclude that no liquefaction is
possible if normalized Cone Penetration Test (CPT) cone resistance, qc, is larger
than 157 tsf (15 MPa). This CPT resistance corresponds to normalized blow
counts between 30 and 60, depending on the grain size of the soil (see Figure 5.2).
If three or more of the above criteria indicate that liquefaction is not likely, the potential for
liquefaction may be considered to be small . If, however, based on the above initial screening
criteria, the potential for liquefaction of a cohesionless soil layer beneath the site of a planned
landfill (new construction or lateral expansion) cannot be dismissed, more rigorous analysis of
liquefaction potential is needed.
75
-------
Liquefaction susceptibility maps, derived on the basis of the some (or all) of the above listed
criteria, are available for many major urban areas in seismic zones (e.g., Kavazanjian et al., 1985;
Tinsley et al., 1985; Hadj-Hamou and Elton, 1988; Hwang and Lee, 1992). However, as most
new MSWLF's are sited outside of major urban areas, these maps may not be available for many
landfill sites. Furthermore, most areas have not been mapped in sufficient detail to be useful for
site-specific studies.
There have been several attempts to establish threshold criteria for values of seismic shaking that
can induce liquefaction (e.g., minimum earthquake magnitude, minimum peak horizontal
acceleration, maximum distance from causative fault). Most of these criteria have eventually been
shown to be misleading, since even low intensity bedrock ground motions from distant
earthquakes can be amplified by local soils to intensity levels strong enough to induce liquefaction,
as observations of liquefaction in the 1985 Mexico City and 1989 Loma Prieta earthquakes
demonstrate.
Most soil deposits known to have liquefied are sand deposits. However, gravel deposits are also
susceptible to liquefaction. Discussion of the liquefaction potential of gravel deposits is beyond
the scope of this document. The reader is referred to Harder (1988) for a discussion of methods
for evaluation of the liquefaction potential of gravels.
5.2 Liquefaction Potential Assessment
Due to the difficulties in obtaining undisturbed representative samples from most liquefiable soil
deposits and to the difficulties and limitations of laboratory testing, the use of in-situ test results
to evaluate liquefaction potential is generally the preferred method for liquefaction potential
assessment among most practicing engineers. Liquefaction potential assessment procedures
involving both the SPT and the CPT are widely used in practice (e.g., Seed and Idriss, 1982;
Ishihara, 1985; Seed and De Alba, 1986; Shibata and Teparaska, 1988). For gravelly soils, the
Becker Hammer penetration test is commonly used to evaluate liquefaction potential (Harder,
1988).
The most common procedure used in practice for the liquefaction potential assessment of sands
and silts, the Simplified Procedure, was originally developed by Seed and Idriss (1982). As used
in engineering practice today, the Simplified Procedure has been progressively revised, extended
and refined (Seed et al., 1983; Seed et al., 1985; Seed and De Alba, 1986; Liao and Whitman,
1986). The Simplified Procedure may be used with both CPT and SPT data. Recent summaries
of the various revisions to the Simplified Procedure are provided by Marcuson et al., (1990) and
by Seed and Harder (1990). Based on the recommendations from these two studies, the Simplified
76
-------
Procedure for evaluating liquefaction potential at the site of a MSWLF can be carried out using
the following steps:
Step 1: From in-situ testing and laboratory index tests, develop a detailed understanding
of site conditions: stratigraphy, layer geometry, material properties and their
variability, and the areal extent of potential problem zones. Establish the most
critical zones to be analyzed and develop simplified sections amenable to analysis.
The data should include location of the water table, either SPT blowcount, N, or
tip resistance of a standard CPT cone, qc, and mean grain size, D50, the unit weight
of the soil, and the percentage of fines (percent by weight passing the No. 200
sieve) for the materials involved in the liquefaction potential assessment.
Step 2: Evaluate the total vertical stress, 0, and vertical effective stress, „', in the deposit
at the time of exploration and for design. Design values should include the
overburden stress due to the landfill. Outside of the waste footprint, the
exploration and design values may be the same if the design ground water level is
at the same elevation as the ground water was during sampling, or they may be
different due to temporal fluctuations in the water table.
Step 3: Evaluate the stress reduction factor, rd. The stress reduction factor is a soil
flexibility factor defined as the ratio of the peak shear stress for the soil column,
(max)d > to mat of a rigid body, (max)r. There are several ways to determine rd. For
depths less than 40 ft (12 m), the average value from Figure 5.3 (Seed and Idriss,
1982) can be used. Alternatively, the following equation proposed by Iwasaki et
al. (1978) can be used:
rd 1 0.015 D (5.1)
where D is depth in meters.
If results of a site response analyses (e.g., a SHAKE analysis) are available, rd can
be determined directly from results of such analysis, as:
I mvx)@depth D
j
*- o'@depth D ^°mm %'©surface
77
-------
where amax is the peak ground surface acceleration and g is the acceleration of
gravity.
Use of the results of a site response analysis to evaluate rd is considered to be
generally more reliable than either of the two simplified approaches and is strongly
recommended for sites that are marginal with respect to liquefaction potential (sites
where the factor of safety for liquefaction is close to 1.0).
Step 4: Calculate the critical stress ratio induced by the design earthquake, CSREQ, as:
CSREQ = 0.65 (a^/g) rd (0/0') (5.3)
Step 5: Evaluate the standardized SPT blowcount, N^. N^ is the standard penetration test
blowcount for a hammer with an efficiency of 60 percent (60 percent of the
nominal SPT energy is delivered to the rods). The "standardized" equipment
corresponding to an efficiency of 60 percent is specified in Table 5.4. If
nonstandard equipment is used, N^ is determined as:
(5-4)
where C60 is the product of various correction factors. Correction factors
recommended by various investigators for some common non-standard SPT
configurations are provided in Table 5.3. Alternatively, if CPT data are used, N^
can be obtained from the chart relating Ngo to qc and D50 shown in Figure 5.2 (Seed
and De Alba, 1986).
Step 6: Calculate the normalized standardized SPT blowcount, (N^. (ty ^ is the
standardized blow count normalized to an effective overburden pressure of 1 tsf
(2000 psf or 950 kPa) in order to eliminate the influence of confining pressure.
The most commonly used way to normalize blowcount is via the correction factor,
CN, shown in Figure 5.4 (Seed et al., 1983). However, the closed-form expression
proposed by Liao and Whitman (1986) may also be used:
CN = (l/or (5.5)
where „' equals the vertical effective stress at the sampling point in tons/ft2.
As illustrated in Figure 5.4, Equation 5.5, and the correction factor curves are
valid only for depths greater than 3 m (10 ft). For depths of less than 3 m (10 ft),
78
-------
Seed et al. (1983) suggested that a correction factor of (CN)3 = 0.75 be applied.
According to these recommendations, the normalized standardized blowcount is
calculated as:
= Na • CN (D 3m) (5.6a)
' N60 • (CN)3 (D < 3 m) (5.6b)
There is some indication that other factors such as grain size distribution can
influence CN (Marcuson and Bieganousky, 1977). However, considering the
uncertainties involved in the Standard Penetration test itself, the above correlations
should be adequate for engineering purposes.
Step 7: Evaluate the critical stress ratio (CSR) at which liquefaction is expected to occur
during an earthquake of magnitude M 7.5 as a function of (N,)^. Use the chart
developed by Seed et al. (1985), shown in Figure 5.5, to find CSR (= av/0').
Step 8: Calculate the corrected critical stress ratio resisting liquefaction, CSRL.
Corrections applied to the CSR calculated in Step 7 include: kM, the correction
factor for magnitudes other than 7.5; k, the correction factor for stress levels larger
than 1 tsf (2000 psf); and k, the correction factor for the driving static shear stress
(this is a correction for non-level ground conditions). CSRL is therefore calculated
as:
CSRL CSR kM k k (5.7)
kM can be determined from chart given in Figure 5.6, developed by interpolation
through tabular data presented by Seed et al., (1983). k can be determined from
the chart presented in Figure 5.7 (Harder, 1988; Hynes, 1988). k depends on the
relative density of the soil, Dr, and can be determined from Figure 5.8, originally
proposed by Seed and modified by Harder, (1988), and Hynes (1988).
Step 9: Calculate the factor of safety against liquefaction, FSL, as:
FSL CSRL I CSRBQ (5.8)
79
-------
There is no general agreement on the appropriate factor of safety against liquefaction (NRC,
1985). There are cases where liquefaction-induced instability has occurred prior to complete
liquefaction, i.e., with a factor of safety greater than 1.0. However, when the design ground
motion is extreme or conservative, most geotechnical engineers are satisfied with a factor of
safety, FSL, greater than or equal to 1.0. It should be noted that the Simplified Procedure is
aimed primarily at moderately strong ground motions (0.2 g < amax < 0.5 g). If the peak
horizontal acceleration is larger than 0.5 g, more sophisticated, truly non-linear effective stress-
based analytical approaches should be considered. Computer programs for non-linear evaluation
of liquefaction potential described in the technical literature include DESRA-2 (Lee and Finn,
1978) and its derivative codes DESRAMOD (Vucetic, 1986) and D-MOD (Matasovi, 1993),
DYNAFLOW (Prevost, 1981), TARA-3 (Finn et al., 1986), LINOS (Bardet, 1987), DYSAC2
(Muraletharan et. al., 1991), and FLAG (Itasca Consulting Group, Inc., 1992).
An example of liquefaction analysis using the Simplified Procedure is presented in Appendix A.
5.3 Liquefaction Impact Assessment
For the soil layers for which the factor of safety against the liquefaction is unsatisfactory, a
liquefaction impact analysis should be conducted. A liquefaction impact analysis may consist of
the following steps:
Step 1: Calculate the magnitude and distribution of liquefaction induced settlement by
multiplying the post-liquefaction volumetric strain, v, by thickness of the liquefiable
layer, H. v can be estimated from chart presented in Figure 5.9 (Tokimatsu and
Seed, 1987). An alternative chart has recently been proposed by Ishihara
(Ishihara, 1993). However, application of Ishihara's chart requires translation of
normalized SPT blowcount (N,)^) values determined in Section 5.1 to Japanese-
standard Nj values (N! = 0.833 (N,^; after Ishihara, 1993). The magnitude of
seismic settlement should be calculated at each boring or CPT sounding location
to evaluate the potential variability in seismic settlement across the site.
Step 2: Estimate the liquefaction-induced lateral displacement, L. The empirical equation
proposed by Hamada et al. (1987) may be used to estimate L in meters:
L 0.75 (//)12 (5)1 (5.9)
in which H is the thickness of the liquefied layer in meters and S is the ground
slope in percent.
80
-------
The Hamada et al. (1987) formula is mainly based on Japanese data on
displacements of very loose sands for soil deposits having a slope, S, less than
10%. Therefore, Equation 5.9 should be assumed to provide only as a rough
estimate of lateral displacement. Since the equation does not reflect the density,
or (N^eo value of the soil, or the depth of the liquefiable layer, it may provide a
conservative estimate of lateral displacement for denser sands or for cases where
the soil liquefies at depth. Note that estimate of lateral displacement by this
equation predicts large liquefaction-induced lateral displacements in areas of
essentially flat ground conditions. More complex methods for assessment of the
potential for lateral spreading are available and can be used where appropriate
(Youd, 1995).
Step 3: In areas of significant ground slope, or in situations when a deep failure surface
may pass through waste and through underlying liquefied layers, a flow slide can
occur following liquefaction. The potential for a flow slide to occur should be
checked using conventional limit equilibrium approach for slope stability analyses
(discussed in Section 6 of this document) together with residual shear strength in
zones in which liquefaction may occur. Residual shear strength can be estimated
from the penetration resistance values of the soil using the chart proposed by Seed
et al. (1988) presented in Figure 5.10. Seed and Harder (1990) and Marcuson et
al. (1990) present a further guidance for performing a post-liquefaction stability
assessment using residual shear strengths.
The above liquefaction-associated deformation phenomena, if too great in magnitude, can
adversely impact the integrity of the landfill containment structures. The question the engineer
must answer is "What magnitude of deformation is excessive?" The magnitude of acceptable
deformation should be determined by the design engineer on a case-by-case basis. Seed and
Bonaparte (1992) report that calculated seismic deformations along the liner-waste interface on
the order of 0.15 to 0.30 m (0.5 to 1.0 ft) are generally deemed to be acceptable in current
practice in California. As cover deformations are readily observable and damage to the cover is
repairable, larger deformations are typically considered acceptable along interfaces in the cover
system than along liner system interfaces. At the current time, determination of allowable
deformations remains a subject requiring considerable engineering judgement.
5.4 Liquefaction Mitigation
81
-------
If the seismic impact analysis presented in Section 5.3 yields unacceptable deformations,
consideration may be given to performing a more sophisticated liquefaction potential assessment
and to liquefaction potential mitigation measures. Generally, the design engineer has the
following options: (1) proceed with a more advanced analysis technique; (2) design the facility
to resist the anticipated deformations; (3) remediate the site to reduce the anticipated deformations
to acceptable levels; or (4) choose an alternative site. These options may require additional
subsurface investigation, advanced laboratory testing, more sophisticated numerical modeling,
and, in rare cases, physical modeling. Discussion of these techniques is beyond the scope of this
study.
Design to resist anticipated deformations could include the use of reinforced earth, structural
walls, or buttress fills keyed into non-1 iquefiable strata to resist the effects of lateral spreading.
A variety of techniques exist to remediate potential liquefiable soils and mitigate the liquefaction
hazard. Table 5.4 presents a summary of available methods for improvement of liquefiable soil
foundation conditions (NRC, 1985). The cost of foundation improvement can vary over an order
of magnitude, depending on site conditions (e.g., adjacent sensitive structures) and the nature and
geometry of the liquefiable soils. Remediation costs can vary from as low as several thousand
dollars per acre for dynamic compaction of shallow layers of clean sands in open areas to upwards
of $100,000. per acre for deep layers of silty soils adjacent to sensitive structures.
5.5 References
Bardet, J.P. (1987), "LINOS, a Nonlinear Finite Element Program for Geomechanics and
Geotechnical Engineering," University of Southern California, Los Angeles.
Dobry, R., Powell, D.J., Yokel, F.Y., and Ladd, R.S. (1980), "Liquefaction Potential of
Saturated Sand - The Stiffness Method," Proc. 7th World Conference on Earthquake Engineering,
Istanbul, Turkey, Vol. 3, pp. 25-32.
Finn, W.D. Liam, Yogendrakumar, M., Yoshida, N. and Yoshida, H. (1986), "TARA-3: A
Program for Nonlinear Static and Dynamic Effective Stress Analysis,", Department of Civil
Engineering, University of British Columbia, Vancouver, British Columbia, Canada.
Hadj-Hamou, T., and Elton, D.J. (1988), "A Liquefaction Potential Map for Charleston, South
Carolina," Report No. GT-88 1, Toluane University, New Orleans, Louisiana, 67 p.
82
-------
Hamada, M. Towhata, I., Yasuda S., and Isoyama, R. (1987), "Study on Permanent Ground
Displacements Induced by Seismic Liquefaction," Computers and Geomechanics 4, pp. 197-220.
Harder, L.F., Jr. (1988)*, "Use of Penetration Tests to Determine the Cyclic Loading Resistance
of Gravelly Soils During Earthquake Shaking," Ph.D. Dissertation, University of California,
Berkeley, California.
Hwang, H. and Lee, C.S. (1992), "Evaluation of Liquefaction Potential in Memphis Area, USA."
Proc. 10th World Conference on Earthquake Engineering, pp. 1457-1460.
Hynes, M.E. (1988)*, "Pore Pressure Generation Characteristics of Gravel Under Undrained
Cyclic Loading," Ph.D. Dissertation, University of California, Berkeley, California.
Ishihara, K. (1985), "Stability on Natural Deposits During Earthquakes," Proc. llth International
Conference on Soil Mechanics and Foundation Engineering, San Francisco, California, Vol. 1,
pp. 321-376.
Ishihara, K., Kokusho, T., and Silver. M.L. (1989), "Recent Developments in Evaluating
Liquefaction Characteristics of Local Soils," State-of-the-Art Report, Proc. 12th International
Conference on Soil Mechanics and Foundation Engineering, Rio de Janeiro, Brazil, Vol. 4,
pp. 2719-2734.
Ishihara (1993), "Liquefaction and Flow Failure During Earthquakes," Geotechnique 43, No. 3,
pp. 351-415.
Itasca Consulting Group (1992), "FLAG, A Finite Difference Computer Program for Two-
Dimensional Static and Dynamic Analysis."
Iwasaki, T., Tatsuoka, F., Tokida, K-I and Yasuda, S. (1978), "A Practical Method for Assessing
Soil Liquefaction Potential Based on Case Studies at Various Sites in Japan," Proc. 2nd
International Conference on Microzonation, San Francisco, California, Vol. 2. pp. 885-896.
Kavazanjian, E. Jr., Roth, R.A., and Echezuria, H. (1985), "Liquefaction Potential Mapping for
San Francisco," Journal of Geotechnical Engineering, ASCE, Vol. Ill, No. 1, pp. 54-76.
Lee, M.K.W., and Finn, W.D.L. (1978), "DESRA-2, Dynamic Effective Stress Response
Analysis of Soil Deposits with Energy Transmitting Boundary Including Assessment of
* Available through University Microfilms International, (313) 761-4700 Ext. 3879, or (800) 521-0600 Ext. 3879
83
-------
Liquefaction Potential," Soil Mechanics Series No. 36, Department of Civil Engineering,
University of British Columbia, Vancouver, Canada, 60 p.
Liao, S.S.C. and Whitman, R.V. (1986), "Overburden Correction Factors for SPT in Sand,"
Journal of Geotechnical Engineering, ASCE, Vol. 112, No. 3, pp. 373-377.
Marcuson, W.F., III, and Bieganousky, W.A. (1977), "Laboratory Standard Penetration Tests
on Fine Sands," Journal of the Geotechnical Engineering Division, ASCE, Vol. 103, No. GT-6,
June, pp. 565-588.
Marcuson, W.F., HI, Hynes, M.E. and Franklin, A.G. (1990), "Evaluation and Use of Residual
Strength in Seismic Safety Analysis of Embankments," Earthquake Spectra, Vol. 6, No. 3,
pp. 529-572.
Matasovi, N. (1993)*, "Seismic Response of Composite Horizontally-Layered Soil Deposits."
Ph.D. Dissertation, Civil and Environmental Engineering Department, University of California,
Los Angeles, 452 p.
Muraleetharan, K.K., Mish, K.D., Yogachandran C., and Arulanandan, K., (1991), "User's
Manual for DYSAC2: Dynamic Soil Analysis Code for 2-Dimensional Problems," Report,
Department of Civil Engineering, University of California, Davis, California.
NRC (1985), "Liquefaction of Soils During Earthquakes." National Research Council, Committee
on Earthquake Engineering, Washington, DC.
Prevost, J.H. (1981), "DYNAFLOW: A nonlinear transient finite element analysis program."
Department of Civil Engineering and Operational Research, Princeton University, (Last update
January 1994).
Riggs, C.O. (1986), "North American Standard Penetration Test Practice," In Use oflnSitu Tests
in Geotechnical Engineering, ASCE Geotechnical Special Publication No. 6, pp. 949-965.
Seed, H.B. and Idriss, I.M., (1982), "Ground Motions and Soil Liquefaction During
Earthquakes," Monograph No. 5, Earthquake Engineering Research Institute, Berkeley,
California, 134 p.
' Available through University Microfilms International, (313) 761-4700 Ext. 3879, or (800) 521-0600 Ext. 3879
84
-------
Seed, H.B., Idriss, I.M. and Arango, I. (1983), "Evaluation of Liquefaction Potential Using
Field Performance Data," Journal of GeotechnicalEngineering, ASCE, Vol. 109, No. 3, pp. 458-
482.
Seed, H.B., Tokimatsu, K., Harder, L.F. and Chung, R.M. (1985), "Influence of SPT
Procedures in Soil Liquefaction Resistance Evaluations," Journal of Geotechnical Engineering,
ASCE, Vol. Ill, No. 12, pp. 1425-1445.
Seed, H.B. and De Alba, P. (1986), "Use of SPT and CPT Tests for Evaluating the Liquefaction
Resistance of Sands." In Use ofln-Situ Tests in Geotechnical Engineering, ASCE Geotechnical
Special Publication No. 6, pp. 281-302.
Seed, H.B., Seed, R.B., Harder, L.F., Jr., and Jong, H.-L. (1988), "Re-Evaluation of the Slide
in the Lower San Francisco Dam in the Earthquake of February 9, 1971." Report No.
UCB/EERC-88/04, University of California, Berkeley, California.
Seed, R.B., and Harder, L.F., Jr. (1990), "SPT-Based Analysis of Cyclic Pore Pressure
Generation and Undrained Residual Strength," Proc. H.B. Seed Memorial Symposium, Vol. 2,
pp. 351-376.
Seed, R.B., and Bonaparte, R. (1992), "Seismic Analysis and Design of Lined Waste Fills:
Current Practice," In Proceedings of Stability and Performance of Slopes and Embankments - II,
Vol. 2, Berkeley, California, ASCE Geotechnical Special Publication No. 31, pp. 1521-1545.
Shibata, T., and Taparaska, W. (1988), "Evaluation of Liquefaction Potentials of Soils Using
Cone Penetration Tests," Soils and Foundations, Vol. 28, No. 2, pp. 49-60.
Skempton, A.W. (1986), "Standard Penetration Test Procedures and the Effects in Sands of
Overburden Pressure, Relative Density, Particle Size, Ageing and Overconsolidation,"
Geotechnique, Vol. 36, No. 3, pp. 425-447.
Tinsley, J.C., Youd, T.L., Perkins, D.M., and Chen, A.T.F. (1985), "Evaluating Liquefaction
Potential," in J.I. Ziony, ed, Evaluating Earthquake Hazards in the Los Angeles Region, an Earth
Science Perspective," U.S. Geological Survey Professional Paper 1360, pp. 263-315.
Tokimatsu, K. and Seed, H.B. (1987), "Evaluation of Settlements in Sands due to Earthquake
Shaking," Journal of Geotechnical Engineering, ASCE, Vol. 113, No. 8, pp. 861-879.
85
-------
Vucetic, M. (1986)*, "Pore Pressure Buildup and Liquefaction of Level Sandy Sites During
Earthquakes," Ph.D. Thesis, Rensselaer Polytechnic Institute, Troy, New York, 616 p.
Youd, T.L., and Perkins, D.M. (1978), "Mapping Liquefaction-Induced Ground Failure
Potential," Journal of Geotechnical Engineering, ASCE, Vol. 104, No. GT4, pp. 433-446.
Youd, T.L. (1995) "Liquefaction-Induced Lateral Ground Displacement," Proceedings of the
Third International Conference on Recent Advances in Geotechnical Earthquake Engineering and
Soil Dynamics, University of Missouri, Rolla, 2-9 April.
' Available through University Microfilms International, (313) 761-4700 Ext. 3879, or (800) 521-0600 Ext. 3879
86
-------
Table 5.1 Estimated Susceptibility of Sedimentary Deposits to Liquefaction During
Strong Seismic Shaking (Youd and Perkins, 1978).
Type of
deposit
0),
General dis-
tribution of
cohesionless
sediments
in deposits
(2)
Likelihood that Cohesionless Sediments,
When Saturated. Would Be Susceptible
to Liquefaction (by Age of Deposit)
<500 yr
(3)
Holocene
(4)
Pleis-
tocene
(5)
Pre-
pleis-
tocene
(6)
(a) Continental Deposits
River channel
Flood plain
Alluvial fan and
plain
Marine terraces
and plains
Delta and fan-
delta
Lacustrine and .,
playa
Colluvium
Talus
Dunes
Loess
Glacial till
Tuff
Tephra
Residual soils
Sebka
Locally variable
Locally variable
Widespread
Widespread
Widespread
Variable
Variable
Widespread
Widespread
Variable
Variable
Rare
Widespread
Rare
Locally variable
Very high
High
Moderate
—
High
High
High
Low
High
High
Low
Low
High
Low
High
High
Moderate
Low
Low
Moderate
Moderate
Moderate
Low
Moderate
High
Low
Low
High
Low
Moderate
Low
Low
Low
Very low
Low
Low
Low
Very low
Low
High
Very low
Very low
?
Very low
Low
Very low
Very low
Very low
Very low
Very low
Very low
Very low
Very low
Very low
Unknown
Very low
Veiy low
?
Very low
Very low
Coastal Zone
Delta
Esturine
Beach
High wave
energy
Low wave
energy
Lagoonal
Fore shore
Widespread
Locally variable
Widespread
Widespread
Locally variable
Locally variable
Very high
High
Moderate
High
High
High
High
Moderate
Low
Moderate
Moderate
Moderate
Low
Low
Very low
Low
Low
Low
Very low
Very low
Very low
Very low
Very low
Very low
(c) Artificial
Uncompacted fill
Compacted fill
Variable
Variable
Very high
Low
—
—
—
87
-------
TABLE 5.2: RECOMMENDED "STANDARDIZED" SPT EQUIPMENT
(After Seed et al., 1985, and Riggs, 1986)
Sampler:
Drill Rods:
Hammer:
Rope:
Borehole:
Drill bit:
Blowcount Rate:
Penetration Resistance
Count:
Standard split-spoon sampler with: (a) O.D. =
2.00 in., and (b) LD. = 1.38 in. (constant - i.e. no
room for liners in the barrel)
A or AW for depths less than 50 ft; N or NW for
greater depths
Standard (safety) hammer with: (a) weight = 140 Ib;
(b) drop = 30 in. (delivers 2,520 in.-lbs which is
60% of theoretical freefall)
Two wraps of rope around the pulley
4 to 5-in. diameter rotary borehole with bentonite mud
for borehole stability (hollow stem augers where SPT
is taken through the stem)
Upward deflection of drilling mud (tricone or baffled
drag bit)
30 to 40 blows per minute
Measured over range of 6 to 18 in. of penetration
into the ground.
Note: If the equipment meets the above specifications, N = Nw and only a
correction for overburden is needed.
88
-------
TABLE 5.3: CORRECTION FACTORS FOR NONSTANDARD SPT PROCEDURE AND EQUIPMENT
CORRECTION FOR
Nonstandard Hammer Type
(DH = donnut hammer; ER = energy ratio)
Nonstandard Hammer Weight or height of fall
/t_l — Vi^jo*tit nf fall in inf^n^c* AA/ — \t/Aifr1if in 1Kc\
Nonstandard Sampler Setup (standard samples with
room for liners, but used without liners)
Nonstandard Sampler Setup (standard sampler with
room for liners, and liners are used)
Change in Rod Length
Nonstandard Borehole Diameter
CORRECTION FACTOR
Cm- = 0.75 for DH w/rope and pully
Cm- « 1.33 for DH^w/trip/auto & ER = 80
H-W
HW 140*30
Cjs = 1.10 for loose sand
Css ** 1.20 for dense sand
CM — 0.90 for loose sand
Css = 0.80 for dense sand
CRL = 0-75 for rod length 0-10 feet
CBD = 1.05 for 6 in. diameter
CBD ~ 1.15 for 8 in. diameter
REFERENCE
• Seed et al. (1985)
calculated per Seed et. al
/•1QO«N
\lyoj)
Seed et al. (1985)
Skempton (1986)
Seed et al. (1983)
Skempton (1986)
00
Note:
= C
ST
'HW
-SS
'RL
'BD
-------
Table 5.4 Improvement Techniques for Liquefiable Soil Foundation Conditions (NRC, 1985).
Method
Principle
Most Swiabk Soil
Conditions/Types
Maximum Effective
Treatment Depth
Economic Sue of
Treated Arc*
Ideal Properties of
Treated Material-
Applications*
Relative
Coses'
(I) Blasting
(2) Vibratory probe
(a) Tcrraprobe
30 m
sometimes; Vib-
towing, 40 m
Cohesiofuess soils
with less than 20ft-
fines.
Can obtain relative
densities to 70-
•0%; may jet vari-
able density; time-
dependent strength
Can obtain relative
densities of 80% or
more. Ineffective in
some sands.
Can obtain high rela-
tive densities (over
85%). food uni-
formity.
Induce KqucJaction in
controlled and lim-
ited states and in-
crease relative den-
sity to potentially
nonliquefUMe
range.
Induce liquefaction in
controlled and lim-
ited states and in-
crease relative den-
sity to potentially
nonliqucfubk
range. Has been
shown effective in
preventing liquefac-
tion.
Induce liquefaction in
controlled and lim-
ited slaves and in-
crease relative
densities to
nonliquenable con-
dition. Is uted ex-
tensively to prevent
liquefaction. The
dense column of
backfill provides (a)
vertical support, (b)
drain* to relieve
pore water pres-
sure, and (c) shear
rcuUnncc in bori-
Z4Mital and inclined
directions. Used to
stabilize slopes and
strengthen potential
failure surfaces or
slip circles.
Low (52.00-
S4.00/m»>
Moderate
(S6.00-
Low to modcr-
ate 20 m > 1 .000 m1
placement of pife partly saturated
volume and by vi- clayey soils; loess.
bration during driv-
ing, increase in lat-
eral effective earth
pressure.
Repeated application Cohcstontcss soils M) m (possibly >3.300 m-
of high-intensity best, other types deeper)
impacts at surface. can also be im-
proved.
Highly viscous grout All soils. Unlimited Small
acts as radial hy-
draulic jack when
pumped in under
high pressure.
Can obtain high den-
sities, good uni-
formity. Relative
densities of more
tnanSO%.
Can obtain high rela-
tive densities, rea-
sonable uniformity.
Re la live densities
of 80% or more.
Grout bulbs within
compressed soil
matrix. Soil mass
as a whole is
strengthened.
Useful in soils with 1
fines. Increase* rd- 2
ativc densities to 3
nonliquefiabk
range. Is used lo
prevent liquefac-
tion. Provides shear
resistance in hori-
zontal and inclined
directions. Useful
to stabilize slopes
and strengthen po-
tential failure sur-
faces or slip circles.
Suitable for some 2
soils with fines; us- 3
able above and be-
low water. In cohc-
sJonlcss soils.
induces liquefaction
in controlled and
limited stages and
increases relative
density to poten-
tially nonliquefiabk
range. Is used to
prevent liquefac-
tion.
Increase in soil rela- 1
five density and 2
horizontal effective 3
stress. Reduce lique-
faction potential.
Stabilize the ground
against movement.
Moderate to
high
Low(VMO-
tt.OO/mM
Low 10 moder-
ate (53.00-
JI5.00/m')
*SP, SW, or SM soils that have average relative density equal to or greater than 85 percent and the minimum relative density not less than 80 percent are in general not susceptible to liquefaction
(TM 5-818-1). D'Appolonia (1970) stated that for soil within the zone of influence and confinement of the structure foundation, the relative density should not be less than 70 percent. Therefore, a criteria
may be used that relative density increase into the 70-90 percent range is in general considered to prevent liquefaction. These properties of treated materials and applications occur only undtr Meat
conditions of soil, moisture, and method application. The methods and properties achieved are not applicable and will not occur in all soils.
* Applications and results of the improvement methods are dependent on: (a) soil profiles, types, and conditions, (b) site conditions, (c) earthquake loading, (d) structure type and condition, aad
(e) material and equipment availability. Combinations of the methods will most likely provide the best and most stable solution.
'Site conditions have been classified into three cases. Case I is for beneath structures. Case 2 is for the not-underwater free field adjacent to a structure, and Case 3 is for the underwater free field
adjacent to a structure.
The costs will vary depending on: (a) site working conditions, location, and environment, (b) the location, area, depth, and volume of soil involved, (c) soil type and properties, (d) materials (sand,
gravel, admixtures, etc.), equipment, and skills available, and (e) environmental impact factors. The costs are average values based on: (a) verbal communication from companies providing the service.
(b) current literature, and (c) literature reported costs updated for inflation.
'A means the method has potential use for Case 3 with special techniques required thai would increase ihe cost.
90
-------
Table 5.4: (continued)
..„._. _. . Mon StuUNe Soil Maximum Effective Economic Size of Ideal Properties of
Method Principle ConAtwos/Type* Treatment Depth Treated Area Treated Materiaf
ADMIXTURE STABILIZATION
< M) Mix-in-placc Lime, cement, or as- Sand, sills, clayl. all >M m 140 m obtained Small Solidified soil piles or
pUcsandw,,,, pW, introduced srf, or loose inor- in Jap«, wall, of relaEvciy
through rotalmg au- game soik. high strength.
ger or special in-
place mixer.
THIiRMAI. .STABII.I/ATION
Ml In-situ vimfica- Mehs soil in place lo All «,,K ,nj ,,vl. >Vlm ua^~n Midif.cd soil piles or
1011 create an obSK),an- wall, of high
like vrtreou, ma.c SIrent!lh ^
n ' vious; more dur-
able than granite or
marble: comprci-
*ive strength, 9-1 1
(CM; &fifi((ing tensile
strength. 1-2 ksi.
Applications4 Case'
Slope subiltzation by I
providing shear re- 2
sisUnce in horuon- 3
lal and inclined di-
rections, which
strengthens poten-
tial failure surface*
or slip circles. A
walj could be used
lo confine an area
of liquefiabk soil so
that liquefied soil
could not flow out
of the area.
Slope stabilization by I
providing shear re- 2
sistancc in hormm- ?
tal and inclined di-
rections, which
strengthens poten-
tial failure surface*
or slip circles. A
wall could be u^cd
lo confine an area
of liquefiabk soil so
that liquefied soil
could not flow out
of the area.
Relative
Costr*
High
(S250.00-
Modrraic
<$VVOO-
(16) Vjbro-repu.cc.
ment stone and
sand columns
lNX>ai:; fine-
bralory equipment) grained soiK.
»OOUm-
All soils
jncrcaxcd vertical and
horizontal load car-
rying capacity.
Density increase in
cohesionless soils.
Shorter drainage
paths.
Reinforced zone of
soil behaves as a
coherent mass.
Provides (a) vertical
support.
-------
0.001
0.01
0.1 1
Grain size (mm)
10
•N
Boundary for most liquefiable soil \ /TsuchidaN
/-—(£>--/ : Boundary for potentially liquefiable soil / * 1970 /
/,t/ '• Tailings slime (Ishihara. 1985)
f f
Liquefied sand in
Chiba-Toho-Oki Eq. (1987)
Liquefied sand in
Niigata Eq. (1964)
/ / . Liquefied sand in
/ / ' Nihonkai-Chubu
Eq.(1983)
100
.SP
'5
-------
e 3
41
C
«!
O.
O Jo«i«olkowtki cl oL (1985)
0 Mufonxxhi and Kobayoshi (1962)
A t*h!horo ond Koga (I96JJ
^ Robertson (19621
V Mitchell (1965}
O Ho«J«» «lolU984)
a Mows/fool
OOI OO2 OO5 O.I O.2 O.5 1
Mean Grain Size. O5O —mm
Figure 5.2 Variation of
1986).
Ratio with Mean Grain Size, D^ (Seed and De Alba,
93
-------
\ roaxjd
0
* \
maxlr
0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 1.0
10
20
30
40
a.
UJ
60
70
80
90
100
I I I I I I I I
AVERAGE VALUES
RANGE FOR DIFFERENT
SOIL PROFILES
I L
Figure 5.3 Stress Reduction Factor, rd (Seed and Idriss, 1982).
94
-------
C 2
v, 3
O.
V
£ *%
c
10
O.2 O.4
O.6
O.8
1.0
1.2
1.4
1.6
qc-volues by CPT-
(Or*4Olo8O%)
-Or«6O to8O%
•Or « 40 to 60%
N-volu«« by SPT
1 I 1
0.2 0.4
0.6 Q8
IX)
1.4
1.6
Figure 5.4 Correction Factor for the Effective Overburden Pressure, CN (Seed et al., 1983).
95
-------
0.6
PERCENT FINES - 35
FINES CONTENT 2 5%
MODIFIED CHINESE CODE
PROPOSAL (CLAY CONTENT - 5%)
NO
LIQUEFACTION
MARGINAL
LIQUEFACTION
LIQUEFACTION
PAN AMERICAN DATA
JAPANESE DATA
CHINESE DATA
60
Figure 5.5 Relationships Between Stress Ratio Causing Liquefaction and
for Sands for M 7.5 Earthquakes (Seed et al., 1985).
Values
96
-------
J..U _
"" 1.5 -_
**.*
MAGNITUDE CORRECTION FACTO*
D O H* H-* )-»• M M-
D b b K-* ro ca ^
1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 ! 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1
\
\
\
A Se<
TTI*
inv
^
\
\
K
\
id et al. (1
erpolated
I
'
983)
i
i
j
i
678
EARTHQUAKE MAGNITUDE (M)
9
Figure 5.6 Curve for Estimation of Magnitude Correction Factor,
1983).
(after Seed et al.,
97
-------
*•
e
X
T£0 AVERAGE CURVE FOR 6AND
FROM SEED (LETTER. DATED 12&M7)
FOLSOU FOUNDATION GRAVEL
LEGEND
FAIRMONT OAM
LAKE ARROWHEAD OAM
UPPER SAN LEANORO OAM
LOWER CAN FERNANDO OAM SHELL
UPPER SAN FERNANDO OAM SHELL
LOS ANGELES CAM SHELL
PERRIS OAM SHELL. RC - #0. •£.
SAROIS OAM SHELL
SAROIS OAM FOUNDATION
THERMAUTO AFTERBAY OAM FOUNDATION
THERMAUTO FORE6AY OAM FOUNDATION
ANTELOPE OAM IMPERVIOUS MATERIAL
ASWAN OAM OUNE SAND
SACRAMENTO RIVER SANO. Of - »«. «0. 7«. 1«C%
MONTEREY « SAND. Or . «0%
REID 8EDFORO SAND. Of - 40.«0*
NEW JERSEY BACKFILL. FttRC.tt*
I I 1 |_
FOLSOM EMBANKMENT GRAVEL
1.0
S.« 4.« 6.C
EFFECnve OVERBURDEN PRESSURE, KSF
Figure 5.7 Curves for Estimation of Correction Factor k,, (Harder 1988, and Hynes
1988, as Quoted in Marcuson et al., 1990)
98
-------
2.S
2.0
1.6
1.0
0.5
Of - 55 - 70%
D, - 35%
NOTE: RANGES NOT PROVIDED FOR
O, - 40% AND 35% DUE TO
LIMITED DATA
I
!
0.1 0.2 0.3
INITIAL STATIC STRESS RATIO, a
0.4
0.5
Figure 5.8 Curves for Estimation of Correction Factor k,, (Harder 1988, and Hynes
1988, as Quoted in Marcuson et al., 1990)
99
-------
O.G
05
0.4-
0.3
0.2
O.I
Volumetric Stroih-%
O54 3 2
1
0
10
20
30
40
50
Figure 5.9 Curves for Estimation of Post-Liquefaction Volumetric Strain using SPT
Data and Cyclic Stress Ratio (Tokimatsu and Seed, 1987).
100
-------
2000
V)
O.
X
z
u
cc
(X
<
LJ
r
V)
O
LU
z
<
a:
o
z
O
-------
SECTION 6
258.14 SEISMIC IMPACT ZONES:
SLOPE STABILITY AND DEFORMATION ANALYSIS
The potentially large accelerations associated with seismic events can induce significant forces that
may lead to permanent deformations within a MSW landfill. These deformations potentially can
lead to impairment of the functions of the containment system. However, reports of significant
seismic-related damage to MSW landfill containment systems are relatively rare. Several studies
dealing with landfill behavior during the 1989 M 7.1 Loma Prieta earthquake report only minor
damage to landfills, even for the landfills located in the epicentral region or founded on relatively
weak San Francisco Bay mud (Orr and Finch, 1990; Buranek and Prasad, 1991; Sharma and
Goyal, 1991; Johnson et al., 1991). Damage was mostly limited to cracking of earthen cover soils
and disruption of surficial piping systems No geomembrane-lined landfills were impacted by the
Loma Prieta event. At least three modern, geosynthetic-lined landfills were impacted during the
1994 M 6.7 Northridge Earthquake. While preliminary studies indicate that no major damage
occurred, the geosynthetic liner system was torn in at least two locations above the limit of waste
placement at one of the landfills (EERC, 1994; (Matasovi, et al., 1995).
Numerous methods and procedures are currently available to evaluate static slope stability
(Duncan, 1992). Most of the methods available are, in some form, suitable for seismic stability
analyses. They can be used in conjunction with several different approaches for seismic analysis,
of which the following two conventional methods represent the current state-of-practice: (1)
pseudo-static factor of safety approach, and (2) permanent seismic deformation approach. Both
of these conventional approaches to seismic stability assessment are based on the principles of
limit equilibrium analysis.
In the pseudo-static factor of safety approach, a seismic coefficient is specified to represent the
effect of the inertial forces imposed by the earthquake upon the potential failure mass and a factor
of safety is defined in the conventional manner as the ratio of the ultimate shear strength of the
slope elements to the maximum shear stresses induced in those elements by seismic and static
loadings. The main drawback of the pseudo-static factor of safety approach lies in its inability
to rationally relate the value of the seismic coefficient to the characteristics of the design
earthquake. Use of the peak acceleration (expressed as a fraction of gravity) as the seismic
coefficient in conjunction with a pseudo-static factor of safety of 1.0 has been shown to give
excessively conservative assessments of slope performance in earthquakes.
In contrast to the pseudo-static factor of safety approach, the permanent seismic deformation
approach involves the calculation of cumulative seismic deformations. The most commonly used
102
-------
method for calculating the permanent seismic deformation of slopes is termed the Newmark
method (Newmark, 1965). In this approach, the potential failure mass is treated as a rigid body
on a yielding base. The acceleration time history of the rigid body is assumed to correspond to
the average acceleration time history of the failure mass. Deformations accumulate when the rigid
body acceleration exceeds its yield acceleration. The yield acceleration is the horizontal
acceleration that results in a factor of safety of 1.0 in a pseudo-static limit equilibrium analysis.
The calculation of permanent seismic deformations using the Newmark approach is depicted in
Figure 6.1. Acceleration pulses in the time history that exceed the yield acceleration are double
integrated to calculate cumulative relative displacement. In a Newmark analysis, relative
displacement is often assumed to accumulate in only one direction, the downslope direction. With
this assumption, the yield acceleration in the other (upslope) direction is implicitly assumed to be
larger than the peak acceleration of the failure mass being analyzed.
In practice, both the pseudo-static factor of safety and permanent seismic deformation approaches
are often combined in a unified seismic slope stability and deformation analysis. Such an analysis
outlined in Section 6.2 of this guidance document.
6.1 Key Material Properties
To perform seismic slope stability analyses, estimates of the unit weight and (dynamic) shear
strength parameters of various components of the landfill are needed. Unfortunately, a large
amount of uncertainty often exists as to appropriate values for some of these parameters.
Evaluation of the material properties of MSW required for a slope stability analysis can be a
difficult task. This is due to the paucity of field and laboratory measurements of MSW properties,
the cost and difficulty in making project specific field measurements, and the heterogeneous nature
of MSW. The following sections summarize the information currently available for estimating
key material properties of MSW.
6.1.1 Unit Weight
Values of unit weight for MSW reported in the literature are summarized in Table 6.1 (Fassett
et al., 1994). Initial values of MSW unit weight can be estimated from landfill gate receipts and
survey elevations of the waste face. Current regulations in California require operators to achieve
an initial density of at least 1,250 Ibs per cubic yard (7.3 kN/m3 or 46 lb/ft3). Average values of
MSW unit weight can also be estimated based upon the total gate receipts over the life of a landfill
and survey data. Average values for MSW unit weight cited by landfill operations and used in
practice for landfill capacity estimates typically vary from 8.6 to 10.2 k/m3 (55 to 65 Ibs/ft)
(Kavazanjian et al., 1995). Landfill-specific values of MSW unit weight will depend upon actual
103
-------
operational practice. For instance, significantly higher MSW unit weights have been reported for
a landfill that used an unusually high percentage of daily cover soil (Richardson and Reynolds,
1991).
The MSW unit weights in Table 6.1 do not account for the increase in density with depth that
occurs in MSW due to its compressibility or to changes that occur with time. Kavazanjian et al.
(1995) have demonstrated that the variation of density with depth can have a significant influence
on the results of static and dynamic stability and seismic response analyses. The dashed line on
Figure 6.2 (Kavazanjian et al., 1995) shows the density-depth relationship developed for one
southern California landfill on the basis of field measurements of density and laboratory
measurements of waste compressibility (Earth Technology, 1988). Based upon the density-depth
profile developed by Earth Technology (1988), the initial and average unit weights cited above,
and representative compressibility values for MSW reported by Fassett et al. (1994), Kavazanjian
et al. (1995) developed the MSW unit weight profile shown by the solid line on Figure 6.2 for use
in stability and seismic response analyses of MSW landfills in the absence of landfill-specific data.
6.1.2 Interface Shear Resistance
The interface shear resistance between geosynthetic components (e.g., a geomembrane and
geotextile interface) and between soil and geosynthetic components (e.g., a geomembrane and low
permeability soil interface) from static laboratory tests are generally used for dynamic stability
analyses. Typical values for peak and residual interface friction angles have been reported by
Williams and Houlihan (1986, 1987), Seed and Boulanger (1990), Koerner (1991), and Byrne
(1994). Byrne (1994) recommended that the interface friction angle used in dynamic analysis be
evaluated on the basis of compatibility between the load-deformation curve from laboratory testing
and the calculated seismic deformation.
Some investigations have reported slight differences between static and dynamic interface
strengths (Kavazanjian et al., 1991). Others have reported that interface shear strengths appear
to be independent of the frequency content and number of cycles of motion (Yegian and Lahlaf,
1992). However, considering the uncertainties inherent to other material properties, it appears
that the shear strength measured in static tests may be reasonably used to represent the dynamic
interface shear strength in seismic stability and deformation analyses.
6.1.3 Low Permeability Soil
The dynamic shear strength of clay soils may be influenced by the amplitude of the cyclic deviator
stress, the number of applied cycles, and the plasticity of the clay (Makdisi and Seed, 1978). In
many cases, the static shear strength may be the same as or even greater than the shear strength
104
-------
for dynamic loading. For saturated, normally consolidated soft clays, the dynamic shear strength
can be assumed to be equal to at least 80% of the static undrained strength with a high degree of
confidence. However, for sensitive and stiff clays, cyclic loading can lead to reductions in
strength and accumulated deformations can exceed the strain at which peak strength is mobilized,
resulting in reduction of strength to residual values.
6.1.4 Granular Soil Shear Strength
The cyclic shear strength of a dry or unsarurated granular soil (sand or gravel with degree of
saturation less than 80 percent) can be assumed to be equal to or greater than the drained ultimate
(large strain) static shear strength. In saturated sands, seismic loading can significantly alter the
dynamic shear strength. Evaluation of the potential for shear strength reduction in a saturated
or almost saturated cohesionless soil (low plasticity silt, sand, or gravel) subject to dynamic
loading may require sophisticated cyclic laboratory testing. Alternatively, a residual shear
strength may be assigned to the soil based upon either undrained laboratory tests or in situ test
results. Evaluation of residual shear strength from laboratory tests is not recommended due to
the difficulties associated with testing. Use of residual strengths derived from in-situ testing is
in general considered more reliable. However, use of residual shear strengths in a pseudo-static
stability assessment can result in a very conservative assessment of the pseudo-static factor of
safety and/or yield acceleration and is not recommended for most problems (Marcuson et al.,
1990).
6.1.5 MSW Shear Strength
The available data on MSW shear strength is relatively limited. Available data includes laboratory
test results on reconstituted samples and strength values backfigured from field load tests and case
histories of landfill performance. Laboratory and field tests have consistently shown shear
strengths in excess of a cohesion of 200 psf (10 kPa) and a friction angle of 20 degrees (Landva
and Clark, 1990 and Richardson and Reynolds, 1991). Table 6.2 presents a compilation of the
available data of MSW developed by GeoSyntec (1993). Table 6.3 presents a compilation of
lower bound friction angles backfigured from observations of the satisfactory performance of steep
side slopes at existing landfills by GeoSyntec (1993) based upon the assumption of a cohesion of
100 psf (5 kPa). Figure 6.3 presents a bi-linear strength envelope for MSW developed by
Kavazanjian et al. (1995) based upon evaluation of the data in Tables 6.2 and 6.3.
Observations of the satisfactory performance of landfill slopes in major earthquakes indicates that
the dynamic shear strength of MSW may be significantly greater than the static shear strength.
Figure 6.4 (Siegel et al., 1990) shows combinations of cohesion, friction angle, and yield
acceleration that resulted a pseudo-static factor of safety of 1.0 in analyses of the slopes of the
105
-------
Oil landfill. . Observations of the satisfactory performance of the Oil landfill slopes in recent
earthquakes combined with this plot indicate that the dynamic shear strength of MSW mobilized
during seismic loading is greater than the values indicated in Figure 6.3.
6.1.6 Sensitivity Studies
It is strongly recommended that all stability analyses of MSW landfills be performed using
parametric studies to clearly identify the sensitivity of the performance of the landfill to the
material properties used in the analysis. If performance depends significantly on a given
parameter, then additional laboratory or field testing may be required to better define appropriate
properties for design.
6.2 Seismic Stability and Deformation Analysis
A prerequisite for performing a seismic slope stability and deformation analysis is performance
of a static slope stability analysis. The seismic stability and deformation analysis is carried out
using the same basic model(s) of landfill (waste mass and foundation) and containment system
used in the static analysis. The following steps are used in a typical seismic slope stability and
deformation analysis:
Step 1: Reinterpret the cross-sections analyzed in the static stability analysis and assign
appropriate dynamic strength parameters. In cases where it is not clear whether
drained or undrained shear strength parameters are appropriate for the dynamic
analysis, follow guidelines presented in Duncan (1992).
Step 2: Evaluate the seismic coefficient, ks. There are many different views on how to
define ks (e.g., Seed and Martin, 1966; Seed, 1979; Marcuson, 1981; Hynes and
Franklin, 1984). The most reasonable definition appears to be one that regards the
seismic coefficient as an empirical factor. This definition recognizes the
limitations of the pseudo-static slope stability analysis in representing the actual
effects of an earthquake on the slope. Unfortunately, this definition provides no
guidance to selection of an appropriate value of ks. Seed (1979) reports that clay
slopes and embankments designed with a minimum pseudo-static factor of safety
of 1.15 using a seismic coefficient of 0.15 have experienced "acceptable"
deformations in earthquakes of magnitude less than 7.5 and intensity less than 0.75
g. However, Seed's definition of acceptable deformations appears to include
deformations of over one meter in some cases.
106
-------
Figure 6.5 shows the results of Newmark seismic deformation analyses performed
by Hynes and Franklin (1984) using 387 strong motion records and 6 artificial
accelerograms. Based upon this data and their experience with seismic response
analyses of slopes and embankments, Hynes and Franklin (1984) concluded that
slopes and embankments with a yield acceleration equal to half the peak ground
acceleration would experience permanent seismic deformations of less than one
meter ( 3 ft) in any earthquake, even for embankments where amplification of
acceleration by a factor of three occurs. In the absence of amplification, or if
amplification is taken into account in determining the peak acceleration, the Hynes
and Franklin data suggest that deformations will remain less than 0.3 m (1 ft) for
yield accelerations less than or equal to one-half the peak acceleration.
Therefore, based upon the work of Hynes and Franklin, it appears that the
maximum value of ks may be determined as ks = 0.5 • amax/g to limit permanent
seismic deformations to less than 0.3 m (1 ft), where amax is peak horizontal
acceleration at the ground surface for analyses of the liner system and at the top of
the landfill for analyses of the cover system. amax can be estimated either using the
simplified methods presented in Section 4 of this guidance document or from the
results of a seismic response analysis.
Step 3: Perform the pseudo-static stability analysis. If the minimum factor of safety, FSmin,
exceeds 1.0 and 0.3 m (1 ft) of deformation is acceptable, the seismic stability
analysis is completed.
Step 4: If the pseudo-static factor of safety is less than 1.0 or the acceptable deformation
is less than 0.3 m (1 ft), perform a Newmark deformation analysis. This is done
with the following three steps:
1) Calculate the yield acceleration, ky. The yield acceleration is usually calculated in
pseudo-static analyses using a trial and error procedure in which the seismic
coefficient is varied until FSmin = 1.0 is obtained. The lowest yield acceleration
for all possible failure surfaces passing through the liner, cover, and/or waste mass
should be evaluated.
2) Calculate the permanent seismic deformation. The permanent seismic deformation
may be calculated using either simplified design charts (e.g., Hynes and Franklin,
1984; Makdisi and Seed, 1978) or a formal time-history analysis in which the
excursions of the average acceleration time history above the yield acceleration are
double integrated.
107
-------
3) Compare the calculated permanent seismic deformation to the allowable maximum
permanent displacement, umax. Seed and Bonaparte (1992) report that umax values
of 0.15 to 0.3 meters (0.5 to 1.0 feet) are typically used in practice for design of
geosynthetic liner systems. For cover systems, where permanent seismic
deformations may be observed in post-earthquake inspections and damage to
components can be repaired, larger permanent deformations may be considered
acceptable. In fact, some regulatory agencies consider seismic deformations of the
landfill cover system primarily a maintenance problem.
Several investigators have presented simplified charts based upon the results of Newmark
deformation analyses for estimating permanent seismic deformations. Makdisi and Seed (1978)
developed the chart shown in Figure 6.6 from the results of two-dimensional finite element
analyses of embankments founded upon rock. This chart includes the effect of amplification of
seismic motions by an embankment and provides upper and lower bounds on the permanent
deformation as a function of magnitude. Hynes and Franklin (1984) developed the chart shown
in Figure 6.5 from classical Newmark "sliding block on a plane" analyses. The Hynes and
Franklin chart does not consider amplification or magnitude effects but includes time histories
which encompass a wide range of soil conditions. Due to the uncertainties in using a simplified
design chart and the characteristics and limitations discussed above, the use of the upper bound
curves from Makdisi and Seed (Figure 6.6) for simplified analysis of the permanent seismic
deformation potential of the waste mass and liner system. The mean + curve from Hynes and
Franklin (Figure 6.5) is recommended for simplified permanent seismic deformation analysis of
the cover system.
If a seismic response analysis has been performed, a formal Newmark seismic deformation
analysis can be performed by using the acceleration or shear stress time histories from the
response analysis. Jibson (1993) describes the analytical procedure for performing such an
analysis. To evaluate the permanent displacement of the landfill mass, the average acceleration
time history of mass above the critical failure plane (the failure surface with the lowest yield
acceleration) should be used. The average acceleration time history may be calculated as the
average of the acceleration time history of each layer above the interface weighted according to
the unit weight and thickness of each layer. Alternatively, the average acceleration time history
may be calculated from the shear stress at the interface divided by the total vertical stress above
the interface, as described by Repetto et al. (1993). To calculate the permanent deformation of
the landfill final cover, either the average acceleration time history of the cover or the shear stress
time history at the cover-waste interface divided by the total vertical stress at the interface should
be used in the Newmark analysis. Particularly for landfills in the eastern United States, where
the earthquake acceleration time history may contain relatively enriched high frequencies, the
108
-------
formal Newmark deformation analyses may yield significantly lower seismic deformations than
simplified Newmark analyses using Figures 6.5 and 6.6.
6.3 Additional Considerations
Stability of the underlying foundation soil is an important consideration in evaluating the overall
performance of the landfill, particularly if a layer (or layers) in the foundation is susceptible to
liquefaction, as illustrated in Figure 6.7. The potential for a liquefaction induced flow failure may
be analyzed using limit equilibrium analyses by employing residual shear strengths in the
potentially liquefiable zones. In this type of post-earthquake stability assessment, the seismic
coefficient should be set equal to zero (Marcuson et al., 1990). If the residual shear strength is
conservatively assessed using minimum values of SPT blow counts (or CPT tip resistance) within
the potentially liquefiable layer(s), a factor of safety of 1.1 may be considered as acceptable.
Evaluation of seismic settlement potential, as described in Section 5.3, still must be conducted to
assess the impact of liquefaction on the landfill.
In some situations, it may be convenient to treat the final cover of the landfill as an infinite slope.
In these situations, the pseudo-static factor of safety and yield acceleration for the cover may be
assessed using the following general equations for the stability of an infinite slope (Matasovi,
1991):
c/( z cos2 ) tan 1 (z d )/( z) * tan
5! » £ (6.1)
/t tan
c/( z cos2 ) tan 1 (z d )l( z) tan
t (6.2)
1 tan tan
where FS = factor of safety, ky = yield acceleration, k,, = seismic coefficient, =unit weight of
slope material(s), w= unit weight of water, c = cohesion, = angle of internal friction of the
assumed failure interface or surface, z = depth to the assumed failure interface or surface, and
dw = depth to the water table (assumed parallel to the slope). The above equations yield the
factor of safety and yield acceleration explicitly for both cohesive (c 0 ) and cohesionless soils (c
= 0). If there is no downslope seepage, the depth to the water table, d^,, should be set equal to
the depth to the assumed failure plane, z.
109
-------
6.4 References
Bray, J.D., Augello, A.J., Leonards, G.A., Repetto, P.C., and Byrne, R.J. (1995), "Seismic
Stability Procedures for Solid Waste Landfills," Journal of Geotechnical Division, ASCE, Vol.
121, No. 2, pp. 139-151.
Buranek, D. and Prasad, S. (1991), "Sanitary Landfill Performance During the Loma Prieta
Earthquake," Proc. 2nd International Conference on Recent Advances in Geotechnical Earthquake
Engineering and Soil Dynamics, St. Louis, Missouri, pp. 1655-1660.
Byrne, R.J. (1994), "Design Issues with Strain-Softening Interfaces in Landfill Liners," Proc.
Waste Tech '94 Landfill Technology Conference, National Solid Waste Management Association,
Charleston, South Carolina, 26 p.
Duncan, J.M. (1992), " State-of-the-Art: Static Stability and Deformation Analysis," Proc.
Stability and Performance of Slopes and Embankments - II, Vol. 1, pp. 222-266.
Earth Technology (1988), "In-Place Stability of Landfill Slopes, Puente Hills Landfill, Los
Angeles, California," Report No. 88-614-1, The Earth Technology Corporation, Long Beach,
California.
EERC (1994), "Preliminary Report on the Seismological and Engineering Aspects of the
January 17, 1994 Northridge Earthquake," Earthquake Engineering Research Center, Report
No. UCB/EERC-94-01, Berkeley, California.
Fassett, J.B., Leonards, G.A. and Repetto, P.C. (1994), "Geotechnical Properties of Municipal
Solid Wastes and Their Use in Landfill Design," Proc. WasteTech 94 - Landfill Technology
Conference, National Solid Waste Management Association, Charleston, South Carolina, 32 p.
GeoSyntec Consultants (1993), "Report of Waste Discharge, Eagle Mountain Landfill and
Recycling Center," Supplemental Volume 1, Report prepared for the Mine Reclamation
Corporation, GeoSyntec Consultants, Huntington Beach, California.
Hynes, M.E. and Franklin, A.G. (1984), "Rationalizing the Seismic Coefficient Method,"
Miscellaneous Paper GL-84-13, U.S. Army Engineer Waterways Experiment Station, Vicksburg,
Mississippi, 34 p.
110
-------
Jibson, R.W. (1993), "Predicting Earthquake-Induced Landslide Displacements using Newmark's
Sliding Block Analysis, Transportation Research Record, 1411, Transportation Research Board,
National Research Council, Washington, D.C., pp. 9-17.
Johnson, M.E., Lew, M., Lundy, J. and Ray, M.E. (1991), "Investigation of Sanitary Landfill
Performance During Strong Ground Motion from the Loma Prieta Earthquake of October 17,
1989," Proc. 2nd International Conference on Recent Advances in Geotechnical Earthquake
Engineering and Soil Dynamics, St. Louis, Missouri, pp. 1701-1708.
Kavazanjian, E., Jr., Hushmand, B., and Martin, G.R. (1991), "Frictional Base Isolation Using
a Layered Soil-Synthetic Liner System," Proc. 3rd U.S. Conference on Lifeline Earthquake
Engineering, Technical Council on Lifeline Earthquake Engineering Monograph No. 4, Los
Angeles, California, pp. 1140-1151.
Kavazanjian, E., Jr., Matasovi, N., Bonaparte, R., and Schmertmann, G.R. (1995), "Evaluation
of MSW Properties for Seismic Analysis," Proceedings of the Geoenvironment 2000 Specialty
Conference, ASCE, Vol. 2, pp. 1126-1141, New Orleans, Louisiana, 24-26 February 1995.
Koerner, R.M. (1991), Designing with Geosynthetics, 2nd Edition, Prentice Hall, Englewood
Cliffs, New Jersey, 652 p.
Landva, A.O. and Clark, J.I., (1990), "Geotechnics of Waste Fill," In Geotechnics of Waste Fill -
Theory and Practice, ASTM STP 1070, pp. 86-103.
Makdisi, F.I., Seed, H.B. (1978), "Simplified Procedure for Estimating Dam and Embankment
Earthquake-Induced Deformations," Journal of Geotechnical Engineering Division, ASCE
Vol. 104, No. GT7, pp. 849-867.
Marcuson, W.F. (1981), "Earth Dams and Stability of Slopes Under Dynamic Loads,"
Moderators Report, Proc. 1st International Conference on Recent Advances in Geotechnical
Earthquake Engineering and Soil Dynamics, St. Louis, Missouri, Vol. 3, pp. 1175.
Marcuson, W.F., HI, Hynes, M.E., and Franklin, A.G. (1990), "Evaluation and Use of Residual
Strength in Seismic Safety Analysis of Embankments," Earthquake Spectra, Vol. 6, No. 3, pp.
529-572.
Ill
-------
Matasovi, N. (1991), "Selection of Method for Seismic Slope Stability Analysis," Proc. 2nd
International Conference on Recent Advances in Geotechnical Earthquake Engineering and Soil
Dynamics, St. Louis, Missouri, Vol. 2, pp. 1057-1062.
Matasovi, N., Kavazanjian, E. Jr., Auguello, A.J., Bray, J.D., and Seed, R.B. (1995) "Solid
Waste Landfill Damage Caused by 17 January 1994 Earthquake," In: Woods, M.C. and Seiple,
R.W., Eds. The Northridge, California, Earthquake of 17 January 1994, California Department
of Conservation, Division of Mines and Geology Special Publication 116, Sacramento, California,
pp. 43-51.
Newmark, N.M. (1965), "Effects of Earthquakes on Dams and Embankments," Geotechnique 15,
No. 2, pp. 139-160.
Orr, W.R. and Finch, M.O. (1990), "Solid Waste Landfill Performance During Loma Prieta
Earthquake," Geotechnics ofWastefills: Theory and Practice, ASTM STP 1070, pp. 22-30.
Repetto, P.C., Bray, J.D., Byrne, R.J. and Augello, A.J. (1993), "Applicability of Wave
Propagation Methods to the Seismic Analysis of Landfills," Proc. Waste Tech '93, Marina Del
Rey, California, pp. 1.50-1.74.
Richardson, G. and Reynolds, D. (1991), "Geosynthetic Consideration in a Landfill on
Compressible Clays," Proc. Geosynthetics '91, Industrial Fabrics Association, Minneapolis,
Minnesota.
Seed, H.B. and Martin, G.R. (1966), "The Seismic Coefficient in Earth Dam Design," Journal
of Soil Mechanics and Foundations Division, ASCE, Vol. 92, No. SM 3, pp. 25-58.
Seed, H.B. (1979), "Considerations in the Earthquake-Resistant Design of Earth and Rockfill
Dams," Geotechnique, Vol. 29, No. 3, pp. 215-263.
Seed, R.B. and Boulanger, R.W. (1991), "Smooth HDPE-Clay Liner Interface Shear Strengths:
Compaction Effects," Journal of Geotechnical Engineering, ASCE, Vol. 117, No. 4,
pp. 686-693.
Seed, R.B., and Bonaparte, R. (1992), "Seismic Analysis and Design of Lined Waste Fills:
Current Practice," Proc. Stability and Performance of Slopes and Embankments - II, Vol. 2,
ASCE Geotechnical Special Publication No. 31, Berkeley, California, pp. 1521-1545.
112
-------
Sharma, H.D. and Goyal, H.K. (1991), "Performance of a Hazardous Waste and Sanitary Landfill
Subjected to Loma Prieta Earthquake," Proc. 2nd International Conference on Recent Advances
in Geotechnical Earthquake Engineering and Soil Dynamics, St. Louis, Missouri, pp. 1717-1725.
Siegel, R.A., Robertson, R.J., and Anderson, D.G. (1990), "Slope Stability Investigations at a
Landfill in Southern California," Geotechnics of Waste Fills-Theory and Practice, ASTM STP
1070, Arvid Landva, G., David Knowles, editors, American Society for Testing and Materials,
Philadelphia, Pennsylvania, pp. 259-284.
Singh, S. and Murphy, B.J. (1990), "Evaluation of the Stability of Sanitary Landfills,"
Geotechnics of Waste Fills - Theory and Practice, ASTM STP 1070, pp. 240-258.
Williams, N.D., and Houlihan, M.F. (1986), "Evaluation of Friction Coefficients Between
Geotextiles, Geomembranes, and Related Products," Proc. Third Conference on Geotextiles,
Vienna, Austria, pp. 891-896.
Williams, N.D., and Houlihan, M.F. (1987), "Evaluation of Interface Friction Properties Between
Geosynthetics and Soils, "Proc. Geosynthetics '87, New Orleans, Louisiana, Vol. 2, pp. 616-627.
P-
Yegian, M.K., and Lahlaf, A.M. (1992), "Dynamic Interface Shear Strength Properties of
Geomembranes and Geotextiles," Journal of Geotechnical Engineering, ASCE, Vol. 118, No. 5,
pp. 760-779.
113
-------
Table 6.1
REFERENCE
Cambell, 1982
Earth Technolgy,
1988
Franklin &
Assoc., 1990
Galante et
at., 1991
Ham, etal.,
1978
Landva et
a!., 1984
Merz & Stone,
1962
Unit Weight Data for MSW (Fassett et al., 19$
LOCATION WASTE WASTE PLACEMENT MOISTUM
TYPE AGE METHOD CONTENT
m
London
Northampton
1
Sussex
Northhampton
1
•
•
•
Los Angeles,
CA
US.
i
S.E. PA
Madison,
Wl
Calgary
Edmonton
•
Mississauga
Red Deer
Vancouver
Waterloo
Winnipeg
Pomona,
CA
CelM
Cell 2
Cell 3
Coll 4
MSW
MSW
MSW
MSW
MSW
MSW
MSW
MSW
MSW
MSW
MSW
79% MSW Fresh
16% Sludge
5% Misc.
MSW Fresh
Soil:MSW
(by vol.)
0:1 Fresh
1:6 Fresh
1:4.6 Fresh
0:1 Fresh
Thin layer*;
Cat 81 6
Sttlitts
Cat 81 «
0.8 ft lifts
LSI D Loader
'Minimal'
Pulverised
'Minimal'
5 ft lifts
Cat 81 6
Thin layers;
Cat 81 8
5 ft lifts
Cat 655 Dozer
30.8
55.3
Cat 828;
8-1 Oft lifts
CatdSOB 38.2
4 ft lifts
(i) 1 18 ft lift; 167.1
wetted to refusal
(il) 4 4 ft lifts; 51.9
Stand. Compact'n
(ilQ tame as (II) 32.5
but no H2O added
(tv) 1 ISttim; 79.5
mln compaction
>4).
• UNIT WEIGHT
IUIAL UHI
(pot) (pet)
83
41
41
29
43
49
40
59
42
80.8 81.8
82.8 53.3
30
83-70
41 30
78-91
84-81
57-«9
88-85
73-42
54-77
88-82
54-73
58.6 22
(Refuse only)
45.9 29.8
36.6 29.1
22.7 12.6S
(Refuse only)
TEST
METHOD
Full-Scat* FWdTett
M
H
It
tl
"
*
II
tt
Plastic Tubes
Estimate
Lined Test Pits
Surveyed
Test PHs
,
Surveyed
Surveyed
Surveyed
Surveyed
RELIABILITY COMMENTS
Fair Do not have details on
measurement technique.
•
•
•
•
•
•
•
•
Good Depth about 50 ft
Poor Estimate based on comp-
onent densities & relative vol.
Very Good
Good Density values are for waste
only; no dally cover Included.
Good Tasting done In 1 983-84 .
Good
Good Compaction Is tub-
standard today
Good
Good
-------
Table 6.1: (continued)
REFERENCE LOCATION WASTE WASTE PLACEMENT MOISTURE
TYPE AQE METHOD CONTENT
(%)
Call 5 1:4.1 Fresh, mixed (v) 1 18 ft lift;
with toll Stand, compact'n
Nalarajan & Bombay, MSW
Rao, 1977 India
OweiS & NJ MSW Fr«sh
Khera, 1966 MSW 'Older'
Pacey, 1982 Mt. View, MSW Fresh
CA
Pfeffer, MSW Fresh I -1. 5 ft lift*
1 992 0-9 Dozer
Richardson & Central MSW 1 -2 yr»
Reynolds, 1991 Maine
Sargunan et Madras. MSW 10-50 Uncompacted
al., 1986 India
Schumaker, 1972 Poor
(From Oweis & Moderate
Khera, 1986) Good
Sharma et Richmond, MSW & ?-40
al., 1990 CA Liquids
Siegel et s CA MSW s-40 yrs
al., 1990
Stone, 1 975 MSW Fresh Stand. Compact'n
No soil no Soil added
Stone & U.S.
Friedland, 1969
41.7
51
47
43
35
41
35
23.6
35.4
53.2
30-46
10-45
29-46
UNIT WEIGHT TEST RELIABILITY COMMENTS
IU1AL
(PC/)
48.5
Av«-64
3.0-15
42
62
70-eo
46
47
44
43
43
44
45
44
49
Ave«96
34-43
16.5
28.8-37
90.6
46
35-46
Ave-4!
S.O.-18
DRY METHOD
fpcO
32.9 Surveyed
Ave«50 'Undisturbed* Samples
S.D-23
Based on pore presi. In tub*
Same
Volume Estimate!
32 Full-Scale Field Test
32
31
32
30
33
34 Full-Scale Field Test
28
22
1 2 Teit Pitt
•In-tltu Tests'f?)
Surveyed
60-1 08 1 3 cm Acrylic Tubes
28-34 Surveyed
National survey
Good
Good
Fair
Poor
Poor
Good
Good
•
Good
Fair
Fair
Fair
Fair
Fair
Poor
6 Sample*
Av* overl 00 ft depth
six 100x100 ft, son
deep Test Cells.
Test reported In 1969;
do not know how dry unit
weight was calculated.
Do not have details on
measurement technique.
MSW to daily cover ratio = 6:1
Samples depths » 15-82 ft
Soil content - 20 to 95%
Results of survey of
103 U.S. Operators.
-------
TABLE 6.2: COMPILATION OF THE AVAILABLE SHEAR STRENGTH DATA ON MSW
SOURCE: GEOSYNTEC (1993)
REFERENCE
Oweis (1985)
Oweis (1985)
Oweis (1985) and
Dvimoff and Munion (1986)
Fagotto and Rimoldi
(1987)
Earth Technology
(1988)
Landva and Clark
(1990)
Singh and Murphy
(1990)
Seigel et al.
(1990)
Richardson and Reynolds
(1991)
TYPE OF DATA
Laboratory tests on baled waste
Field load test at landfill in southern
California
Back calculation based or. failure of a
landfill foundation on marsh clays and
silts
Back calculation from the results of
plate bearing tests
Small-scale triaxial tests on wastes
from LACSD Puente Hills landfill
Laboratory direct shear tests on
municipal solid waste
Laboratory tests performed by others
Small-scale direct shear tests on wastes
from on
Large direct shear tests performed
in situ
RESULTS
m 15" to 25°
c - 1400psf(67kPa)
= 10* and c = 1000 psf (48 kPa)
to
0 = 26' and c = 0
~4
- 10° and c » 500 psf (24 kPa)
to
4> = 30" and c = 0
0 « 22" and c = 600 psf (29 kPa)
= 24" and c = 2,600 psf (124 kPa)
to
= 35"
- 24" and c = 450 psf (22 kPa)
to
= 39" and c = 400 psf (19 kPa)
« =» 0 and c = 700 psf (34 kPa)
to
4> = 42" and c = 0
= 39" to 4> = 53°
* = 18* to 43 "and
c - 200 psf (10 kPa)
COMMENTS
No data on waste types, density, or test methods
is provided. Results correspond to a limiting
strain of 15 to 20%.
Values represent lower-bound strengths as the
slope did not fail. Results are for relatively low
normal stresses. Assumed unit weight of
MSWis45pcf(7kN/mJ).
Values unreliable due to uncertainty in the back
analysis and strain incompatibility between waste
and foundation.
No data on waste types, test procedures, or
test results are provided.
Strength with cohesive component associated with
cover soil dominant waste; coheskraless strength
for refuse dominated waste.
Normal stresses up to about 10,000 psf
(480 kPa). Shear box size was approximately
11 in. (275 mm) by 17 in. (425 mm). Lower
value corresponds to shredded waste, not used.
Strength values bracket a range of results
compiled from work of various researchers and
consulting organizations.
Confining pressures from 2,000 to 12,000 psf
(100 to 570 kPa). Lower value of * is based
upon conservative interpretation.
Normal stresses from 200 to 800 psf (10 to
40 kPa). Unit weight of waste and cover soil
estimated to be 96 pcf (IS kN/nr1).
-------
TABLE 6.3: LOWER BOUND FRICTION ANGLES BACKFIGURED
FROM OBSERVATIONS OF STEEP LANDFILL SLOPES
SOURCE: Kavazanjian et al., (1995)
LANDFILL
(Location)
Lopez Canyon
(California)
Operating Industries, Inc.
(California)
Town of Babylon
(New York)
Private Facility
(Ohio)
AVERAGE SLOPE
Height,
ft(m)
400
(120)
250
(75)
90
(30)
130
(40)
Angle,
H:V
2.5:1
2:1
1.9:1
2:1
MAXIMUM SLOPE
Height,
ft(m)
120
(35)
75
(20)
45
(10)
35
(10)
Angle,
H:V
1.7:1
1.6:1
1.25:1
1.2:1
WASTE STRENGTH
(<£, c = 100 psf)
FS ~ 1.0
25°
28°
30°
30e
FS « 1.1
27«
30°
34°
34°
FS = 1.2
29"
34°
38°
37°
Notes: (1) FS = Assumed factor of safety for back-analysis to estimate <£. Back-analyses were performed assuming c = 100 psf (5 kPa).
(2) Data for Lopez Canyon, Town of Babylon, and Private Facility Landfills obtained from GeoSyntec Consultants project files. Data for Operating
Industries, Inc. Landfill obtained from Siegel et al. (1990).
-------
One-Way Sliding
Only
Positive Direction
Time
D
vfl(t)
Time 13
Yield
Acceleration,.
2
-------
EH
fX,
w
Q
0
25-
50-
75-
100
This Study
Earth Technology (1988)
Extreme Values (Fassett et al., 1994)
Range of Typical Average Values
Derived from Fassett et al. (1994)
0
6 9 12
UNIT WEIGHT (kN/m3)
15
Figure 6.2 Unit Weight Profile for MSW (Kavazanjian et al., 1995)
119
-------
250
£200
CO 1 <=in
CO 10U
PS
100
W 50
CO
0
+ Richardson & Reynolds (1991)
O Lopez Canyon
Operating Industries (OH)
Toim of Babylon B
Private Facility in Ohio
Fagotto & Rimoldi (1987)
Landva & Clark (1990)
0
A
O
24 kPa
0
I I I I I I I I
I I I
50 100 150 200 250 300 350
NORMAL STRESS (kPa)
Figure 6.3 Bi-Linear Shear Strength Envelope for MSW (Kavazanjian et al., 1995)
120
-------
Friction Angle, 0 = 64
Whittier Narrows
earthquake,
Oct. 1,1987
= 5.9
= 0.47 g (peak)
(Garvey Reservoir)
Pasadena earthquake
Dec. 3,1988
= 5.0
kh = 0.22g (peak)
(Bedrock)
kh = 0.10g (peak)
(Top of Landfill)
k, represents horizontal ground acceleration.
180
Cohesion (kPa)
Figure 6.4 Yield Acceleration as a Function of Shear Strength Parameters for the OH
Landfill (Siegel et al.t 1990).
121
-------
1OOO
0.1
1.0
Yield Acceleration / Maximum Acceleration
Figure 6.5 Hynes and Franklin Permanent Seismic Displacement Chart (Hynes and
Franklin, 1984).
122
-------
lOOOj
O.I -
O.OI
U
M
Yield Acceleration
Maximum average acceleration
Displacement (cm)
Richter magnitude
0.2
0.6
0.8
i.O
Figure 6.6 Makdisi and Seed Permanent Displacement Chart (Makdisi and Seed,
1978).
123
-------
"LOCAL FAILURE"
'DEEP FAILURE
AILURE
SAND OR
T r
^ r t r r r t r
r r i
Figure 6.1 Modes of Instability of a MSW Landfill.
-------
APPENDIX A LIQUEFACTION POTENTIAL
EXAMPLE 1 - MSWLF on Level Site - Initial Screening
EXAMPLE 2 - MSWLF on Level Site - Global Stability
125
-------
APPENDIX A EXAMPLE PROBLEM - LIQUEFACTION POTENTIAL
EXAMPLE 1 : This example evaluates die potential for liquefaction at die site of a proposed
above grade MSWLF. The subgrade is composed of fine sands and die water
table is near surface. The analysis neglects die additional normal stresses that
will result from die landfill itself. This is conservative.
— -
i j
1 4 ' &^_A A. * ' 1 A- J • A *: j
J ..) M T *°»4 ^Wa*
1 ! Mis 11
±'
_ . 4cc
> i ! • . '• .
i ' * i < i
! i
Ml ' 1 "*
,~— i j. :-- —4-- • -- — *-
nil i *
Mi sfrv'n i i
; j ! ! i , ' 1 5$ *5
0 i i <0 i "» £
0
i
_vl
?
5
IP
-*
1
*l"
3
IK
-:
5
-:
_
! !
1» ! 54
-t- - ^)^rj' < •• ' K
i f-^{ nsr- | Ifl 0
^j ««Lie»fr 10
i \ ' **
/f j £O^IC6«C^
> ' i I j -. .-
":<.!!! 17 15
^Xf! ! !
| i
Cfl^Lu.a
MKalnlft*
SuJ
»rpUT<:fl!
aoES..
\
*** &C
Log of
Equipm
tElcvatk
^
•
;>-; i A* 20-
; f ! i
!/ 33 25
i i * i !
-T~ 30-
i i'-;aLt UTCb^etJ^
• :: :: i j :: • : ;
III MM M
:• : - - :' • :•
: -: i
' '; >^V "; : r : '!
; i : i i i i i i
x
^
X
.
i<
^
X,
:: • ' i
• •• •
!_, •
1
• * *•*«"
•:->v
j : <
i | j
^C^ci-loKl< £>.«5-7,ft'
APHHJ s i
i i . ! J
e PMM6: 4- «F= e^
M i • • I
i i !
i < i i i !
i : : I ! i '
Boring G-3 >
Ww MODnd B*o 1
w -HO ft Ifci/e 01/23/93
UQHT BROWN CALCAACOUS SAND WITH M.T
(SP-SM). v*ry d«nio lo mMlVirn 4«n««, HIM.
mowt. «om« ih«H«. .
¥ W«t« kv«l «t 64 1 Mt, 1 4:30. 1/29/93.
i
uo
UT nAAtftfld f*U J^AMVMl* «A&«V J*IH
n» mwfnn v^MfAncuvs Bfw DTI^
n*Aiiw 4«w« to o^M. mo^iufti to «o*n«.
Htur«t«(L
8«eom«« Bght 0r«v-brovvn.
End of boring at 26.5 tot.
^z.rx ^ ^ "
? ; i
•; J i
^t> ?E>e w^ti
I .: 1
i 4 *aPT^^o I I !
| i •S*TU«UTJE-C> | Ml
M MM! !
i i ' i i i s i
"^~^^T^"r"-*^ •• -\:^-*-yr1*- ••{-- -V^- - (fQ g •- -• •-;•
i i i ; i ; i i > * *
24-20
"Cftte 3-1
SEISMIC DESIGN OF MUNICIPAL SOLID WASTE LANDFILLS Pgj^ oW
126
-------
APPENDIX A EXAMPLE PROBLEM - LIQUEFACTION POTENTIAL
EXAMPLE 1: The resistance of the existing soil strata is evaluated using the Simplified
Procedure as outlined in Section S. On this page the CSR capacity of the
existing subgrade is calculated.
SEISMIC DESIGN OF MUNICIPAL SOLID WASTE LANDFILLS
127
-------
APPENDIX A EXAMPLE PROBLEM - LIQUEFACTION POTENTIAL
EXAMPLE 1: The CSR generated by the design earthquake is evaluated. The acceleration at
the surface of die site is estimated using die general amplification/attenuation
relationships from Figure 4.2.
j«e:r
Cfc
tKcoeyctgon
, j^vcyey«y
SCfffl
m
•, ,t
. i •
•4—^
16
- ..
2L -1
Ct>»atfu.'
C&
&
—J
> ! i
f 5 ;
n i
! !
^iji k
i^y-T n
i i
O-llff i j :_
0-I12J i I
3
SEISMIC DESIGN OF MUNICIPAL SOLID WASTE LANDFILLS
128
-------
APPENDIX A EXAMPLE PROBLEM - LIQUEFACTION POTENTIAL
EXAMPLE 1: The Factor of Safety against liquefaction calculated for die two critical depths
is greater than 1.0, so the site has limited liquefaction potential.
SEISMIC DESIGN OF MUNICIPAL SOLID WASTE LANDFILLS
PgA of±
129
-------
APPENDIX A EXAMPLE PROBLEM - LIQUEFACTION POTENTIAL
EXAMPLE 2: A potentially liquefiable sand layer is located approximately 30 feet below the landfill. Analyses
have shown that in the case of liquefaction neither the integrity of overlying soil will be
disrupted nor excessive settlement will occur. However, there is a concern that global stability
of the landfill might be disrupted, as shown hi the enclosed diagram. Evaluate the post-
earthquake global stability of the landfill assuming mat the underlying sand layer has liquefied.
!
!
j
-f
i
i
4
i
j
.. _ . i_ .
- • — j —
— h
..4.
~T
— L
— j-
\
i
..-.- HI-II
|
•:
M ! I •= » J i ? ! !
j 1
1 I 1 /
•Mi /
.J — 1 ...» - • — i. ...; ; ( v>
1 i i i ; i \ *
1 1 ! i i ! \
_i_ 1 _j y
-I ill A
\ jfc-p | ; \
E £3 j
J |e j ^
1 |= ^ J ( |
-^^l ' 5
-r i|| —
O S! ;
1 ' g ;
> • "Z.
o g
T . . T- i^
2 Z
5Ci
: UJ <
I/>CO
M MM °*") wiiY
-------
ISHISMIC DESIGN OF MUNICIPAL S
0
c/:
5
r
$
r*
..
,..,,,
i L,,L .
1
! 1 1
1.
1 •
f 1 .IT r ; |--T--, T r T -p- T
The residual shear strength for the liquefied soil was esti
between corrected SPT blow counts and residual shear stre
and reiterated by Marcuson et al. (1990). Based on averaj
layer (liquefied zone), the relationship indicated lower bourn
parameters for the underlying sand layer were therefore tak<
psf for the post-earthquake stability assessment. Slope stal
strength yielded a factor of safety of 1.1 which was deeme
L: J ( ( LJ LI
.,..,„
,...,.
1
2000
IA
Q,
^ 1600
J.
H
C
u.
.... ,. . ..... £
t-
V
a.
<
u.
J 3
i If
c
U.
2
<
C
—
<
C
i;
«
*
trt
UJ
a
|
I !
j t
j
I 1
1
>oo
JOO
100
: 1
\
\
• CI«TMOUA«C - inooet» IK
S'l DATA AHO KttlOOAl 1
O CA*THOUA«t -IKOOttO Lit
5>f 6ATA AHO MSIOUAk 1
O COH1TDUCTION - INtUetO
', *rf^
) 4
EQUIVAl
**
•HMd
.Elf
8
r CLEA
^.
H SI
\
1
rftCMCtH p
nicr*ctioM
tTMCHQTM *
LI6UCFAC1K
f;
Y
i
IND
,
2
SP
i
.....,.,| |«...|., .,..,j.
mated on the bas
ngth for sands de
;e corrected SPT
i residual strength
;n as friction angl
)ility analysis per)
d acceptable.
!. [ 1
1
1-
SO
vel
bio
of
eof
orn
a
f th
opet
w a
200
zer
led
AMO SLIOINC cAst Mtsfoftict IM^HC
AMO ILMMM: CASC Hisromtt «Mt*c
•AUtTCn HAvc BCCN CITlHATCD. "
3N AND SLID4MC CAfC HISTOttlCS.
^-I0«t« JAK f t«»ANOO OAH
16
T BLOWCOUN
j
20
T, (N,
>eo
2
'1
L ^
i t i
e empirical relations!
1 by Seed et al. (198
>unt of 12 for the lot
psf. The shear strenj
o and a cohesion of 2
using the residual sh<
1
L
*••• -^
4 •-
—
—
—
IE
....}.._
i
~ :~
-t—
—-: -
—
,..
~~
........
...„.
I-
iip
8),
>se
?th
00
;ar
—
.._.
••~~
—
,..,.,.
J-
—
—
—
.......
..,,,.
._
......
—
"*****
"""
•~~
-------
APPENDIX B SEISMIC SLOPE STABILITY
EXAMPLE 1 - MSWLF on Ridge
EXAMPLE 2 - MSWLF Above Grade
EXAMPLE 3 - MSWLF on Soft Subgrade
EXAMPLE 4 - MSWLF Displacement Analysis
132
-------
APPENDIX B EXAMPLE PROBLEM - SEISMIC SLOPE STABILITY
EXAMPLE 1: This example reviews die seismic stability of a MSWLF placed on a ridge. The
subgnde stratigraphy at me site has approximately 10-ft of soil profile over me rock
forming the ridge. Two impoituit assumptions ate made at mis point: 1-me peak
bedrock acceleration acts on me bottom of me liner, and 2- due to me ridge, me
seismic cpeffictant is assumed to equal the peak bedrock acceleration.
i ) ;
H--4-4-*
...1 L-U.L
• ': i !
tfcol crap
SEISMIC UbblCiJN Uf MUJNlUUfAL SULliJ WA51H LAJNlJf 1LL5
133
-------
APPENDIX B
EXAMPLE 1:
!
\
~
e;
j
*u
i -
*J-
•
jl^*i/
jut-
j»
:
MH
t
— i-^
|
I
1
i
1
]
]
— i
<
i
!
i
EXA
!
pel «4*V
i
i
!
! -
i
_._,-. .
i
!
i
r-*8*
!
'
M
!
i--
i ^*-*
!Zi^—
j 1
-fi£^-
•
Mi
It
CO*
M.
«o
nil
»LEPI
esetsm
>^fip^n
SWdue
n«tf«| jo
>grade
= i
_
•^
i
i
:
i ! ! 1
'•-'.•
; " j -.
; ^
"^TA
fc\
{
!•>
a*
&TK«t
. . ..._
te^&e,
i
I
i^
S£*Mtf^*
KO
ice
ts.1
to
iah/i
dKM
jl
BU
taba
[he
DM
as \
ildi
T <
F^
k
F3
AV^XT
|
!
EM-SEI
STABLmo
ligherctren
nU confirm
lot be disco
| j
*/(j
-r-
—
] Oo
i
i
)
!
dud
i ' 1 i i i i S
'ROFILE
EXAMPLE #1 RIDGE TOP SITE
} 5
). 1818. 50. 1820. 2
>0. 1820. 850. 2050. 1
J50. 2050. 1000. 2050. 1
LOGO. 2050. 1250. 1980. 1
L250. 1980. 1450. 1980. 2
JO. 1820. 350. 1870. 2
350. 1870. 900. 1945. 2
300. 1945. 1250. 1980. 2
30IL
2
<
-1
J
]
'
55. 65. 200. 20. 0. 0. 0
L20. 120. 500. 25. 0. 0.
EQUAKE ,~ — -,
3.11 0. 0. CCMV a \
BLOCK ^~^Z/
25 2 50.
50. 1820. 350. 1870. .0
351. 1870. 900. 1945. .0
SEISMIC DESIGN
0
lllN
:
i
- I
t
j
SM!
nliiM
deli
gths
uUS
unte
[C
xll
• «
of
dv
! i
Wr
,
SL
aodi
atup
(he
OPE STABHH
ifl is evahiated for
to force the fauun
nflftrrnl •**•!« m*A &>
^:^i «
•HHI"-"! f
riftoutanat
!
;
^iVeis-i
i
f ^
J****^
:
-<••• j
!
i
—*
w/o
V-
2
!
I
:
LINER
> i
i
j
i
*l .zc*
M -
G
u
U) '
\^
\)
\J
r
IS
(C>
V
ui^ds-
VI
>l
! •
- :;
—
J »v>
..i. ,
3,
p-*1^ J
i /
/
— ' !
ota
|TBS
T
a lange of seisouc
> suxfMoe to bottom in the
«geometiy. White*
^^^^^ *
i
~1 i
- "^^^i^^
-
1*
; •<
r*i
i
we
i !
: !
)
:• •;
i * 1 -
:- •,
... — , —
\
\ :
', ',
i i
i «:
I
t
:<
1
i
1
!
i i
• "' ^^
jTCiiLs:
t >
1 «-t M
:-
i_
•h
' i
I I'fc
I \ &***&£*'•$»*
OF MUNICIPAL SOLID
i i :
i
-^=^
. — • —
|
t i
T"7
j i i
J i j '
fi
i
•/
1 \
\ -f
^_j LI
i
t
j
*
•4 — —
;
S
j
t
]
i
^
I
' |
1 1 '
__^S^
i
1^
Te?Hr
i
t
•>-. "'"'
i t i
»U>!
! i
I
i 1 i i
FS» V5Z
WASTE LANDFILLS Pg^ ot^
134
-------
APPENDIX B EXAMPLE PROBLEM - SEISMIC SLOPE STABILITY
EXAMPLE 1: The STABL model was modified to auow the addition of 'sou" types to represent the
liner system. In me example shown, two liner types ate included so mat textured and
smooth sheet could be incorporated at the same time.The shear strength of the liner
'soils' should be obtained from interface friction testing of the liner interfaces.
&
j
/*Uu><
j
i
2»o«r-
I
fS».
•»
! 9
~ PROFILI
EXAMPLI
-11 5
-0. isie
50. 181
850. 2(
:
_
LOOO. :
L250. ]
50. 18:
350. 1J
900. 1<
! i i 1 MM i i ! j ^ i •
i i • j 5 • 1 } i } '» 1 ill
tei vk^ksiil=" ^vjs|&\tUw vj f utvifeei -wvheiA 1 !
i I i i M ' M i M ^
M i '! Ill MM ; !
= i ! ! s ! i ! i ; 3 i z
:•"•'/ > J :• ' -•' •- *•
i M ! i i i ; J^<-^: ! r^
MM ^-f^rn j*?ii—~-r-~z~~~~' r
^^^~*~**^~' — !~~^ ' ; ' :: '
^f^^ J ^ M i ; . ; i i i
Mi! i ! *)*• i • : ; i M "^
: i i i 1 ! ! i ' • : ' !
,^ ^ : ! ; 1 . . : ;
"TA^>^— 3-MV^o.T »"»ij£ ' • ;
^ . ;
#1 RIDGE TOP Sllr, W/ LilNtK
j ^0, 1820. '?- i J^L iV : i >l
JO. 850. 2050. 1 M | \ 1 j I
)50. 1000. 2050. 1 3 j i \ 1 '
J050. 1250. 1980. 1 j* f 5 " ] \ j '• ' ''"
L980. 1450. 1980. 2 if ;\j t
L -7 » O JZJ*J« AO/U* TT *W /J C ~ ' j \^
370. 900. 1945. 3* $ * \
345. 1250. 1979.5 44 *i U ^^s_
50. 1819. 350. 1869.5 2 _J 7' ; ^
-
-
350. 1
900. 1
SOIL
4
65. 65
B69.5 900. 1944.5 2 j» ; , , i
^44.5 1250. 1979.5 2 j0^ t^ ( J^ ,j^
FXc.Trf
,— —
-
X
1 -:
_ i !
5 i
.'TT
-^J-r?
i 1 i
''. ;
i «t
1 i :
; .:
-: j
•i ,
( i
i i ;
1 i i
i
;
j -;
i
t ;
i ;
; ;:
i :
i : |
K> ; i-jg
TY
tf^^j &4 tJ AVl OW "t
L>4A«Wirc3
ue r«:
-------
APPENDIX B EXAMPLE PROBLEM - SE1
EXAMPLE 2: This landfill is built upon a i
gnrfam^ att*»ftlfeniinn ran V**. •
approximately level so that t
peak ground surface accclen
1 ! !
|^
1
I i
i
i i
•w^s/^
S
TJC,^ |
— 1 **'"^
|
I
i
: j
; { 1
Tl"
i 1
j i
i i
1 1
j ; "
C£S ' | S "
* * !
*
44..
1 ! j
! ! 1
> ' ::
i J
5 ' '' MATERIAL ENGINEERING PROPERTIES
Item
NYSDOT
No. 2 Stone
Solid Waste
Soil Line i
Upper Clayey Till
Glaciolacustrine Clay
Sand and Silt
Basal Till
Moist Unit Satu
Weight Unit V
(pcf) (p
rated
Veight
110 110
65 75
125 130
135 138
120 120
130 138
13S 140
Total Stress Analysis
(Short Term)
Cohesion
(psf)
0
300
1000
1800
350-1500*
0
0
Friction Angle
(Degrees)
30*
33'
0
0
0
32"
34'
Effective Stress Analysis
(Long Term)
Cohesion
-------
APPENDIX B EXAMPLE PROBLEM - SEISMIC SLOPE STABILITY
EXAMPLE 2: The STABL model used in this example is much more detailed than (hat used in
Example 1. Each layer of the liner is included in me model. The model can be
simplified by replacing the multiple layer liner with a single layer having the lowest
interface friction properties. This complete model allows evaluation of the influence of
pore water (or kachate) pressures within individual layers on me overall stability.
PROFILE
-EXAMPLE #2 CLAY FOUNDATION
.21 6
0. 35. 119.5 35. 5
'119.5 35. 158.5 48. 5
158.5 48. 163.9 49.8 4
163.9 49.8 164.5 50. 3
164.5 50. 392.5 126. 2
850. 126. 2
3
3
850. 25.1 3
163.9 49.8 174.5 49.8 4
174.5 49.8 230.5 21.8 4
230.5 21.8 850. 24.9 4
158.5 48. 174.5 48. 5
174.5 48. 222.5 24. 5
222.5 24. 232.5 19. 6
232.5 19. 850. 22.1 6
222.5 24. 6
246.5 17. 7
392.5 126.
164.5 50. 174.5 50.
174.5 50. 230.5 22.
230.5 22.
0. 24.
0. 17.
SOIL
8
110. 110. 0. 30.
65. 75. 300. 33
0.
0.
0. 0
0. 0
.1 .1 .0 10. 0. 0.
1000. 0. 0. 0.
1800. 0. 0. 0.
600. 0. 0
0. 32. 0.
0
34. 0
0.
0.
246.5 17. 850. 16. 7
0. 14. 450. 12. 8
450. 12. 850. 10.5 8
125. 130. 1000. 0. 0. 0. 0: I
135. 138. 1800. 0. 0. 0. 0!—T
120. 120. 600. 0. 0. 0. 0 I—f-
130. 138
135. 140
EQUAKE
0.25 0. 0.
LIMITS
3 3 !
0. 8. 300. 8.
300. 8. 500. 6. ;
500. 6. 850. 4. i.
BLOCK !_
50 3 40. ;
174.5 49.95 174.5 49.95 0 .f
-230.5 21.9 230.5 21.9 0. 1
325.0 22.37 425. 22.87 .8
SEISMIC DESIGN OF MUNICIPAL SOLID WASTE LANDFILLS
137
-------
APPENDIX B EXAMPLE PROBLEM - SEISMIC SLOPE STABILITY
EXAMPLE 3: The global seismic analysis shows that die governing failure during an earthquake is
the result of a global failure. Increasing the global factor of safety may require
reduction of the landfill exterior slope or reduction of the waste height
ob
I
"
Ju,
frfr-
U*>
M i 1 ; |
1 /*^ J y ^ ' 1
j ^jiivii >i4 f j f ^^^^*i GA*W *QU>id
f i 1 j
i 1 :
t i i !
IS. M
', ' i. —
i : * r
j ! j -^
" l ^i^ I
'• } ; _L_._^T 1 1
i j i ~ PsJ 1
i ^C1 fcx- : '• * 'x^_ ^ ?
t ,V_t^^| -. ; ; ; :-
• •: : ; T •'
' > i i i
i i '; 1 ]
i KJWtnJ Ciki i j
! «rj S i S
i ] 1 | ! j j i
M I ! Mi
ill] ill
i 1 ! f ! f3 < i
! i i ! ! i :.C^
° ':!':, • *«« i
I ; .: ; 1 M M 1 i
; ) ! i i 1
,;£-_*. 0.1 W- ^.^g^
; -; !
1 i M,,
Jt ^-^^ ' ;
~~^\r\ \ \
\ \ ':
1 ! '
<*tf
i
- -| — -
1
' • ;S\(i6»4t^J'llS
i i i o : i
"4^vo
i i
1
- j l ,
1 I 1
j l
i I
---
—
jcfteif&fXJse
' *<*> i i }
> - f
; , . v
i i : : * i •
'-'.'; ] -.
r~r
! • M •
! i i
i i
| ill
•
f *C
i i } i ! i M
1 i • • '•'.'' \
• t 5 ,
: . 5 S « t
i \ 0 } I
« i j
j_ !
:—
:
„ i
l«««n
I
1
i
•-XB
te
i 1 !
w*
*i
• - • . '::'•':!•'
- - • :".'". •:.'-- - - .- • :: • * • - *,•• -
MATERIAL ENGINEERING PROPERTIES
Moist Unit Saturated
Total Stress Analysis
Cohesion Friction Angle
(psf) (Degrees)
^An»/\£»ww.«*woVs/*jrrc \pO — *2^. -Zp-i-c' «J5 4oo o
L»»cre. -So £0
MtT; ; . M : ; J
O »0
*Hr«»t_ Stress Analysis
(long Term) »
Cohesion Friction Angle
(psO (Degrees)
*%_& o ^ o
\^ G* t2 O
/Voo o
*T»o o
o \o
i ; ^ " '!- i > 5 ? ;
* ~i i i
u^csziz^
SEISMIC DESIGN OF MUNICIPAL SOLID WASTE LANDFILLS PgJ^ of^
138
-------
APPENDIX B EXAMPLE PROBLEM - SEISMIC SLOPE STABILITY
EXAMPLES: This landfill is sited on a (nick deposit of normally consolidated marine clays. The
static slope stability must evaluated the impact of me rate of waste placement, e.g. the
marine clays increase in strength as they consolidate under the weight of the MSWLF.
j ! j i ! ! i 1 i M ! i i
[Sf»rt\£J^si^5$?£i CJr^Cfevi.vf'-f j } } S
; i ; i i i ' i i i i
1 !}: j j | |
jill: ^i^ C'<.' — • \ 'ML \^\<*'
1 i i ! j i M 1 ** 1 T" ^
\£*KJS. 4 '»> VMrtird •**+*&***& STAjeieo TB 4c
j 1 < j "\3flG»»Jf*&IS \afSi tW4^>C(fe.O-ll*>4^5 ^.CAi1-
1 ; i j i \ i ! i ; M* i
in i | f5 ! f T i M 'S1"1
1 MM i U4^[ i i i
M Mil ^*^\ M '• '• '•- :
ki Kli M MM MS
: '• ^ i j j 1 ' ' i ! ! i i
i s ! MsJ i ! ^A^
j I M M^i-rzhr - M -
1 1 111 ) "" i .
I I I i > I ! 'M
,^>*»»">^'>>t,>^'\"\>:»\ A, »,»,,>,,,,,
1 — 1| j , j ' •' • \ i i ' i 1 -i i ; , i .
PROFILE •. ' , '• :
EXAMPLE #3 STATIC STABILITY 1 j i"S(A^
93 .
o. 100. 160. 100. 3 ; SOIL
— 160. 100. 325. 160. 1 MB
— 325. 160. 500. 160. 1 : 60 60 200_ -
160. 100. 182. 82. 2 cr> cn n in
'Mi1
i i '
; j
i
eEt-nnqi^uW^ ! ! !
i Ml "
i_*wc> c««^.pci«^e
rl 'p^i- U?>
! • "^ hs*.
! i : 5 i Cff-
^5; Ml T'
! i !:M
! i/ ! 1
M /; i
i i /i M
1 IxT 1 i 1
> " !, i ! i
^^
i
i
i !
i
M! ! j \
>U IJAnvk fn-fe !
SO. 0. 0. 0
r\ r\ /t
__182. 82. 500. 82. 2 _100. 100. 1000. 0. 0 0 0
160. 99. 182. 81. 3 90 90.<4ol^0. 0 0 o"
182. 81. 500. 81. 3 \ go_ 90 . 40o7\0 . 0. 0. 0
. 80. 500. 80. 4 ; LIMITS /
0 40 500 40 5 -^ i T L
\ ''••'• • . • •: i '' i : ''• f) ft C f\ f\ f\
M M ^ Mi. ^ ._J. M 1 CIRCLE
; : ; i 20 10
" M M , .. \ M \ J120. 300. 0. 500., 0. 10,. 0. 0.
FS*.M4
PS-V4*>
-
SEISMIC DESIGN OF MUNICIPAL SOLID WASTE LANDFILLS Pg2- of±_
139
-------
APPENDIX B EXAMPLE PROBLEM - SEISMIC SLOPE STABILITY
EXAMPLE 3 The static analysis indicates (hat the critical global failure surface passes through the
soft marine clays and is significantly influenced by the rate of waste placement. This
means that the waste must be placed at a rate dial will allow the consolidation
generated pore water pressures to dissipate.
-r-M-t
SOIL
5
60. 60. 200. 30. 0. 0. 0
50. 50. 0. 10. 0. 0. 0
100. 100. 1000. 0. 0. 0. 0
90. 90. 400. 0
90. 90.
EQUAKE
0.20 0.
LIMITS
1 1
0. 100.
BLOCK
50 3 10.
80. 100. 160. 100. 0.
160.5 99.5 182. 81.5 0.
182.5 81.5 500. 81.5 0.
SEISMIC DESIGN OF MUNICIPAL SOLID WASTE LANDFILLS
140
-------
APPENDIX B EXAMPLE PROBLEM - SEISMIC SLOPE STABILITY
EXAMPLE 3 : The pseudo-static seismic analysis indicates a block sliding factor-of-safety,FS, of 1 . 1
assuming 4 - 30 degrees in the MSW. The influence of me assumed MSW internal
angle of friction on the FS is shown below. The assumed MSW properties can
influence the FS as significantly as the seismic coefficient.
\ \ !
&*«
1 i
tj *
! i '
! i
t j
i •- ;'
\ }
! s
i i
; i
i i
s ?
i <
PROFIL
— T7YJVTUTDT
OA/iRUrlJ
9 3
0. 100
MM
MvL ^UoPe *"
i j i {
^f i i
_ .... . ... .j
i i
\ i
.. — . — f _^**^ •
.%l_
X
\ \ ^ ,
i 1 ;
i i ;
: 5 ':
i-
_ ,„ ., ,, 1- ,..i— . I
M L
&X»*««-\-lV f-
1 i
E«»-».»L £o&f*
1 1 i
"f " 'f
; ^
i .
^^
; r
• !
r — >^^ ^ ?
• ; i
.-; i i i
<^vL«!e.«4i- f^
: i i ;
; J i :-
i ? i ;
i ! j :
L_ : i ; s
j-|.^
i • '
__j_^-*'
-:
E
E #3 Seismic Analysis - Longte
. 160. 100. 3
160. 100. 325. 160. 1 !
325. 160. 500. 160. 1 SOIL
-160. 100. 182. 82. 2 5
-182. 82. 500. 82. 2 60. 60. 200.
160. 99. 182. 81. 3 50. 50. 0. 1(
182. 81. 500. 81. 3 10(
0. 80. 500. 80. 4 90
~0. 40. 500. 40. 5 90
; . . : o.:
i > 1 i > ': i , LI1
M ! M ; i • ; i
i i : i ; : i •• •< o .
:. ': : -: :!._.".'- r**n
i" i i J ! j . CIJ
.. i : : M '. : 12
! j 1
i^e^ !
| |
-*=i i 1
3 i
i i *
i i
; ', i
! i
! I ;
p-Sua^UJr-j,
^>-4^ i \ *$/
! ! _/
: * /!
/I 1 i
: i t <
' * -*'fS
1 ! i
inn "i — |— j —
< f i
| : |
30. 0. 0.
3. 0. 0. 0
3. 100. 1000. 0. 0.
. 90. 400. 0. 0. 0.
. 90. 750. 0. 0. 0.
JAKE
20 0 . 0 .
yiiTS
L
0. 500. 0. 0.
RCLE
10
0. 300. 0. 500. 0. 1
i ; " .- ' 1 :• f *. ;
1 i !
! 1 !
Mil
i i M
4 i _ ] _
f\ I j
/!
! !
•• } i
! ! j
i ! I
1 '' '
i i i
^ L. I ,....^.
i ! T
i ' ':
• : ' f
?
o 5-
0. 0 !
0 "~
0
-:
1—
0 . 0 . 0 . :
P«.54*u:<
'
rs-o-4<
K4^r
^
SEISMIC DESIGN OF MUNICIPAL SOLID WASTE LANDFILLS Pg 4 of
141
-------
APPENDIX B EXAMPLE PROBLEM - SEISMIC SLOPE STABILITY
EXAMPLE 4: MSWLF Displacement Analysis
Background
An MSW landfill is founded on 40 m (130 ft) of silt and clay soil with an average shear wave
velocity of 200 m/s (660 ft/s). The landfill has a geomembrane in the base liner and cover.
The weak layer in the base liner is anticipated to be at the filter geotextile / operations layer
interface. The weak layer in the cover is anticipated to be at the filter geotextile / vegetative
layer interface. Stability analyses show yield acceleration (horizontal acceleration for FS = 1.0)
equal to k, = 0.10 g for base liner and It, = 0.07 g for the cover. Algermissen map (USGS
Map MF-2120) shows the MHA for a hypothetical bedrock outcrop at the site equals 0.12 g.
The regional seismicity source zone map (Fig. 3.2) reveals the maximum earthquake magnitude
to be M 6.4.
SEISMIC DESIGN OF MUNICIPAL SOLID WASTE LANDFILLS
Pgj_ of_2_
142
-------
APPENDIX B EXAMPLE PROBLEM - SEISMIC SLOPE STABILITY
EXAMPLE 4: MSWLF Displacement Analysis
Liner Stability Evaluation
Step 1 Find the free-field peak ground acceleration (PGA) at the ground surface using Seed and Idriss
(1982) charts.
From Fig. 4.4b (or from Fig. 4.5), for MHA = 0.12 g, PGA « 0.23 g.
_ Step 2 Compare PGA to k,.
kyis less than PGA. Therefore, a deformation analysis is required.
— Step 3 Estimate liner deformation (i.e., the maximum permanent displacement, u^), as function of PGA -
and ky using Makdisi and Seed (1978) charts.
Using upper bound of the M 6.5 area in Fig. 6.6, for ky / PGA = 0.41 and M 6.5, a value for u,,,,,
of approximately 20 cm (8 in.) is estimated.
Step 4 Determine if the estimated u,^ value is acceptable.
The landfill is an area fill with essentially a flat base and no penetrations through liner. The -
weakest interface (interface for which kj, is calculated) is between the operations layer and -
geotextile. Therefore, based upon engineering judgement, the calculated deformation of 20 cm (8
in.) is considered acceptable.
- Cover Stability Evaluation
Step 1 Find the maximum acceleration (a^J at the top of the landfill using the Idriss (1990) chart.
Using the recommended median relationship from Fig. 4.4b (or from Fig. 4.5) and the PGA of 0.23
g from Step 1 of the liner stability evaluation, an a,,^ value of approximately 0.32 g is estimated
at the top of the landfill.
Step 2 Compare a^ to k,,.
Since k, for cover is less than 0.5 • a,,^, a seismic deformation analysis is required.
Step 3 Estimate cover deformation as function of a^and ky using Hynes and Franklin (1984) charts.
Using mean + a curve from Fig. 6.5, for k>, / a,^ = 0.22, the u^ value of approximately 20 cm
(8 in.) is estimated.
Step 4 Determine if estimated u^ value is acceptable.
The critical (lowest yield acceleration) surface in the cover is anticipated at the filter geotextile /
vegetative layer interface on the side slopes. Since all vertical gas wells that penetrate the composite
cover are on the horizontal deck, the estimated cover deformation of 20 cm (8 in.) is considered
acceptable.
SEISMIC DESIGN OF MUNICIPAL SOLID WASTE LANDFILLS P ofz
143
•U.S. GOVERNMENT PRINTING OFFICE: 1 995-6 50-006 / 2 20 1 6
-------
United States
Environmental Protection Agency
Center for Environmental Research information
Cincinnati, OH 45268
Official Business
Penalty for Private Use
$300
Please make all necessary changes on the below label,
detach or copy, and return 10 the address in the upper
left-hand corner.
if you do not wish to receive these reports CHECK HERE Q;
detach, or copy this cover and retu"1 to the address ;- 'he
upper left-hand corner.
BULK RATE
POSTAGE & FEES PAID
EPA
PERMIT No. G-35
EPA/600/R-95/051
------- |